Conference on Structural Wonders 23 November 2012, Singapore

ENGINEERING THE STRUCTURAL WONDERS OF THE UAE

THE DIFC LIGHTHOUSE, DUBAI

Matthew J. Esther

WS Atkins & Partners Overseas, PO Box 5620 Dubai UAE matthewesther@hotmail.com / matthew.esther@atkinsglobal.com

ABSTRACT

Churchill's saying that 'We shape our buildings, thereafter they shape us' has never rung more true than in our time, when every major city is striving to claim their place in the world and with their identity defined by the building of larger and taller monuments. The design and construction of skyscrapers is a field of engineering that is continuously fuelled by human endeavour and daring, with the pursuit of building higher constantly pushing the boundaries of material performance, mirroring the pace of technological progress in other engineering fields.

Today’s tall buildings form a vital part of metropolitan living and are the backbone of successful urban developments: this requires innovative engineering solutions to ensure they are efficient and sustainable and achieve the necessary levels of comfort and safety for the communities they serve. Innovations in the design of super-tall structures has seen a trend evolving away from single material solutions, with the optimum design often being a concrete / steel hybrid. In a number of recent projects Atkins has made use of long span steel structures to pair concrete core walls which engage these primary structural elements to act as 'hybrid mega-frames'[2,3,4]. These have the advantage of enhancing lateral stability whilst creating large column free volumes and bringing simplicity to the potentially complex structural arrangements often associated with such bold and ambitious architecture.

The boom experienced in Dubai in the decade prior to the financial crisis saw advances in computer based technology which has enabled creative and daring engineering solutions to meet the challenges of not only building taller, but increasingly slender buildings. This paper showcases the process behind the design of a truly unique structure and how the use of technology combined with an engineered approach from first principles resolved a complex structural arrangement into a relatively simple and buildable solution.

This paper aims to give an insight into the approach to tall building design with particular reference to practices in the UAE. This project is currently on hold and at the time of writing this paper the design and peer review were complete, the building permit was issued and the foundations built. The author feels that despite this hiatus, the innovations and lessons learnt from the project deserve to be shared with the engineering community.

KEYWORDS Super-tall buildings, slender structures, wind and seismic design, hybrid mega-frame, coupled core walls, vibration, progressive collapse, construction sequence analysis, LEED accreditation and environmental aspects, Dubai ground conditions. ABOUT THE COMPANY Atkins was established by Sir William Atkins in 1938 and has since grown to become the largest engineering consultancy group in Europe, employing over 18,500 staff who operate from 200 offices in 150 countries. Since opening its first office in the Middle East in 1976 the company's project portfolio has expanded from the more traditional civil engineering type projects to multi-disciplinary infrastructure, transportation and building developments which have in many ways defined the identity and skylines of cities in the Gulf. In the world of tall buildings this evolution is represented in the design of more specialised structures with a step change that began in 1994 with the commission for the design of the Burj Al Arab - a building which has become the yard-stick against which subsequent generations of young architects and engineers at Atkins measure their designs. In the decade since the completion of the Burj Al Arab, Atkins has developed a portfolio of over 45 completed tall buildings in the Middle East including Almas Tower, The Address Hotel and the Bahrain World trade Centre.

Burj Al Arab The Address Hotel Almas Tower Bahrain WTC

INTRODUCTION

The site of the Dubai International Financial Centre (DIFC) Lighthouse is located in Dubai's financial district, a 110 acre free-zone of the same name, located at the heart of this modern city which harbours ambitions of becoming a financial powerhouse located at the crossroads of Europe and Asia. This design of this building was in response to the Client’s desire for an expressive and timeless building which would embody the dynamism of an upcoming international financial centre and the energy of an ambitious young nation.

The elegant lines of this building mirror that of the other rectilinear icons in the DIFC precinct particularly the Emirates Towers and DIFC Gate (Figs. 1 and 2): it adopts the cross-bracing motif which is so prominent in the DIFC Gate building and extrudes it vertically. The primary structural form is expressed in the elevations and incorporates the DIFC logo within the diamond shape formed by the criss-crossing diagonals over the height of this elegantly tapering tower. Its styling cues are reminiscent of the Bank of China building in Hong Kong and the Hancock building in Chicago, but beyond the bold expression of diagonals on elevation, these similarities end: the mega-braces do not only act to laterally brace the building but also form part of a large gravity structure that transfers loads to the two core walls creating a 40mx30m column free void 5 storeys high at the podium level of the tower.

The site for this 402m tower is on a 12,200 m2 plot and is set to have a built up area of 173,000m2 with 90,000m2 of grade A office space on 64 floors, a visitor’s centre, over five basements, a podium level and lower floors of the tower which accommodate a leisure deck with health club, swimming pool, executive lounge and business centre. The building is crowned at level 67 with a 123m high steel lattice tower which features a viewing deck offering uninterrupted views of the bustling city.

Figures 1 and 2: The DIFC Lighthouse depicted next to the DIFC Gate Building

and Emirates Towers

ENVIRONMENTAL CREDENTIALS The design aspires to set the standard for low-carbon commercial buildings locally and aims to considerably reduce its total energy consumption through a number of design provisions which include passive solar architecture, low energy, low water engineering solutions, recovery strategies for both energy and water. The initial concept was for the upper portion of the tower to accommodate three horizontal-axis wind turbines to harness wind to generate a portion of the building's power in a manner similar to that of the Bahrain World Trade Centre, another Atkins designed building. However, further studies found that the marked differences between the two sites and building shapes had an impact on the potential for energy generation and as a result the idea was shelved. In 2009 the project received LEED Platinum Rating pre-certification from the US Green Building Council.

BUILDING DESCRIPTION At a height of 402m this building is considered super-tall and with a slenderness ratio of 13.4, or the ratio of the total building height above ground to narrowest base dimension, it is very slender. This resulted in key parts of the design focussed on its aerodynamic response under wind load and the associated occupant comfort at the uppermost levels and the viewing deck. The tower footprint occupies approximately one quarter of the site and is located just off-centre of what is a trapezoidal shaped plot. The tower, podium and basement are integral and are not separated by a movement joint which does give rise to large in-plane forces at the upper basement level which result from the horizontal reactions of the tower and thermal stresses. These drawbacks were far outweighed by the reduced effective building height and the added strength afforded by engaging the basement structure which lowers the overturning effects on the foundations.

At its base the tower has dimensions of 88 x 30m and it tapers linearly in both elevations to a plan dimension of 46.5m x 8m at the viewing deck level with the narrowest occupied floor at level 66 having a width of 15.6m. The narrow sides taper 2.96° whilst the broad face of the building tapers by 1.43° over the full height of the building (Figs. 3 and 4). The framing consists principally of a pair of closed reinforced concrete core walls which measure 24 x 30m and sit 40m apart and are paired from level 8 up to the viewing deck level with 3 planes of steel mega-braces set out conveniently at 45° and tied at every level by the floor beams (Fig. 5). The central row of mega-braces run from level 8 to level 67 and is vertical whilst the facade pair of mega-braces follow the taper of the building from level 8 to the viewing deck level.

SOIL CONDITIONS The soils throughout Dubai generally consist of a layer of fine sand overlying a weakly cemented sedimentary rock strata with the water table usually located at between 2 to 4m below the ground level, but which in parts of the emirate can be as much as 10 to 15 m bgl. For the project site the soil investigation found 4 to 8m of fine to medium fine sands (Aeolian Sands) sitting on a gravelly calcareous sandstone with a rock strength classification of weak (or a UCS of up to 5MN/m2). A total of 10No boreholes were drilled to depths of 40m, 70m and 100m which represents roughly 50%, 125% and 200% the width of the raft below the invert level of the raft which is accepted as a rule of thumb as being sufficiently accurate

estimate of the size of the stress bulb in the underlying strata to estimate the foundation settlement.

The selection of piled foundations in tall buildings in the UAE is the standard approach to foundation solutions, although the concept of the piled raft has gained more acceptance from regulatory authorities in recent years. Projects such as the Burj Khalifa and other tall buildings of note have demonstrated the benefits of a piled raft given the high bearing capacities routinely achievable in the sandstone layers located at depths of 8 to 12m.

Figures 4 and 5: Front and Side Elevations

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Narrow Face Tapers 2.96o

Broad Face Tapers 1.43o

Viewing Deck

Free-formed 'blob' Structure

Figure 5: A 3D segment from levels 8 to 12 showing general building arrangement

ENABLING WORKS AND BASEMENT The enabling works for this building were substantial given the depth of excavation required to create the five-level basement, this was exacerbated by the complexity of construction works on all adjacent plots, which in the most extreme case meant that the foundations on one flank were being undercut by some 14m by our planned reduced dig level. The typical depth of excavation of 19m below ground level (bgl) was required for the podium structure with the tower raft requiring a further 3m in depth. The retaining scheme was tailored to suit each boundary condition and consisted of secant piled walls with one or two levels of tie backs, which after detailed analysis ended up being threaded between the piled foundations of the adjacent buildings.

The soil investigation recorded the water table at approximately 12 - 15m bgl which is unrealistically low and was caused by a number of dewatering schemes which were running in adjacent sites and which caused the simultaneous draw down of the water table and skewed the readings from the investigation. Older records were consulted and experience of neighbouring sites was relied upon to establish a water table level at 2.0m bgl. This gave rise to a design pressure head of 16m which required the retaining walls to be 600mm thick typically at basement 5 level and which step in incrementally at every floor to 350mm. Provisions were also made in these basement walls for knock out panels to allow the car park to be linked at each level to those of adjacent buildings.

The site was dewatered with a draw down to a minimum of 500mm below the lowest reduced dig level which would have to be maintained until such times as the hydrostatic pressure was balanced by the weight of the building which in this case was when the podium slab level is cast. There are a number of piles under the podium designed to be in tension in the permanent condition.

Floor plates between three planes of mega-trusses that span between the concrete cores

Steel mega-trusses

Concrete 'punched' core walls

Secondary stair & lift cores

Thickenings at corners of core walls

FOUNDATIONS The foundation schemes for the project can be divided distinctly into the piles supporting the tower and those carrying the podium. The tower is founded on a pair of 3.4m thick rafts each measuring approximately 33 x 38m sitting on 1.5m diameter x 43m long bored piles arranged at a minimum spacing of 2.5 pile diameters and clustered essentially in rows of threes along the perimeter of the core wall where it connects to the raft. This ensures that loads are transferred to the piles in the most direct route, thus reducing the flexural stresses in the raft. These piles are grade C60 concrete (fcu - cubic strength of concrete) with a design capacity of 27500kN which is limited by the codified guidelines which place an upper limit of approximately 0.25fcu on the design strength of the concrete. The required durability was achieved by using a triple blend concrete design mix of OPC (Ordinary Portland Cement) with 70 to 85% GGBFS (Ground Granulated Blast-Furnace Slag) replacement and 8% microsilica.

The podium foundations consist essentially of a uniform 1.4m thick piled raft with piles grouped at column and wall locations with additional piles arranged on an intermediate grid. These piles were 750 and 900mm diameter x 12 m long CFA piles which act essentially as rock anchors as most are in tension in the permanent condition. CFA piles were chosen for the shorter piles as these did not need to be installed under a bentonite slurry and as a result were faster to install and more cost effective than the bored piled alternative.

The raft was designed for the worse of the load effects arising from the code specified gravity and transient loads with the most critical portion of raft being at the interface of the tower raft which tapers to connect to the thinner podium raft. The loading sequence during construction would be such that prior to switching off the dewatering when construction is at podium slab level, all piles would be in compression but with the steady build-up of hydrostatic pressures the forces in the podium piles would turn from compression to tension forces. With the simultaneous construction of the tower, the flexural stresses in the raft would increase with the maximum moments (6EIδ/L2) increasing in proportion to the predominantly elastic settlements with a commensurate increase in shear forces. The brute force approach was adopted and the portion linking the two types of raft was designed for maximum credible differential movement as this was thought to be more prudent an option when compared to plugging a joint subject to hydrostatic pressures of 160kN/m2.

To ensure the highest standards of construction the authorities in Dubai enforce a pile testing regime which requires that one sacrificial test pile be provided for each pile diameter selected and tested to 200% working loads, 1% of all working piles be statically load tested (with Kentledge or reaction piles) and 5% of working piles be dynamically tested and that 10% of working piles be cross-sonic core integrity tested. In addition to this, the capacity of the 132 piles were validated by testing one working pile to 150% of its design load with acceptance being of settlements less than 1% of the pile diameter at working loads and 25mm at the full test load.

DEVELOPMENT OF STRUCTURAL FRAME The structural solution for the tower evolved from a number of options which were developed to meet the client's brief and architectural intent with the final choice being between a steel intensive option and a steel/concrete hybrid solution. The structure to latticed tower forming the viewing deck at levels +279m to +402m and the 40m long office floor plates and transfer structure between the cores were common to both schemes, with the principal difference

being the construction of the cores where the difference between a steel lattice frame and a punched concrete core wall were considered. When considering the merits of each option a number of factors were considered, notably: economy and speed of construction, availability resources and materials locally, lead-in time for a total steel solution, delivery and fabrication, inherent fire resistance, acoustic performance, flexibility for building services requirements, robustness and quite importantly for a tall slender building the inherent damping of each option. The market conditions prevalent at the time were of rocketing steel prices and shortages in the steel market resulting in long order times and transportation bottlenecks for larger steel sections which immediately favoured the hybrid solution. The key deciding factor was finally cost where the concrete option in a like for like comparison proved to be a quarter of the cost of the steel intensive solution.

GRAVITY SYSTEM The RC core walls are linked by 3 planes of steel mega-braces stacked 6 high over the height of the office floors to level +279m with 3 more bays extending up to the viewing deck level. Each mega-brace module is 10 floors high (40m) spanning 40m between the only vertical faces in the core walls and are connected at levels 8, 17, 27, 36, 46, 56A, and 65 via a fabricated node (consisting of a large stiffened gusset plate) welded to an embedded steel column cast into a thickening in the core walls. These mega-braces triangulate with columns and floor beams which they intersect with at 8.0m centres vertically and 4.0m centres horizontally respectively to transfer gravity loads back to the core walls. Under gravity loads the upper portion of the mega-braces are in tension and subsequently pick up load at every column intersection turning into large compression forces at the lower portion of the diagonal. The vertical resultant forces at the base of the diagonals are transferred to the core walls and the net horizontal shear force distributed between the core walls and primary floor beams by virtue of their relative stiffness. A similar effect was found in the columns, where the upper portion acts to pull down the diagonals, and was primarily due in part to the effects of switch-on gravity in the modelling approach used and also by virtue of forces induced by horizontal and vertical compatibility at the intersection of the diagonals, spandrel beams and columns. This was subsequently refined by a more accurate construction sequence analysis towards the end of the design phase, but in the interim, the connection of the columns to the underside of the braces were modelled as longitudinally slotted. This leads to a more accurate split of the gravity loads within each module inscribed by a mega-brace into a series of large chevrons (Fig. 6) and gives a more accurate assessment of the forces in the columns which would otherwise end up being undersized due to the tension initially seen in the upper portion of the columns. This approach also assisted us in making a preliminary estimate of the forces in the diagonals arising from the shortening of the concrete cores from elastic, shrinkage and creep effects in a process that was refined throughout the design process, well ahead of when the modelling of the long term effects of the core walls was carried out.

Figure 6: Isolation of load into chevron shaped modules

The spandrel beams located within the diamonds (the area formed within the braces - a term coined by I. M. Pei when describing the expression of the structure in the facade of the Bank of China Building in Hong Kong)[25] are connected rigidly to the columns to reduce the effect of bending stresses in the presence of large axial loads. These axial forces arise as the horizontal resultant of the axial forces in the braces and by the horizontal and vertical compatibility at the intersection of beam column and diagonal as described above. The nature and magnitude of this horizontal force varies depending on its location (internal or external to the diamond), its position relative to the connecting node (central or at supports) and are very sensitive to the assumed stiffness of the core walls. The nature of the forces within the diamonds is essentially tensile under predominantly gravity loads, increasing to a maximum at mid-height of the diamond or at the level where the braces connect to the core walls (Fig. 7). There are also large compression forces occurring in the spandrel beams immediately above and below the central nodes which differs slightly from the behaviour described in Smith and Coull [5] where only the first spandrel beam is expected to be in tension, this phenomenon is explained by the constraint provided by the stiff concrete cores.

The lower diagonals connect to the concrete core walls immediately below the plant room floor at level 8 where the net horizontal shear force is the resultant of the large axial forces of the order of 32MN and 45MN in the external and internal diagonals respectively. This horizontal force is split between the concrete core walls and the tie beam at that level and is distributed in proportion to their relative stiffness with the forces on the core walls, giving rise to permanent locked-in flexural and shear stresses whilst the forces in the cross ties are balanced by virtue of symmetry by the corresponding diagonal connected to the other core wall at the same level. This force is split between the cores and cross-ties with the distribution being time dependent and the load-share increasing in the tie due to the 'softening' of the core walls due to creep arising from a large sustained lateral load. Predicting the behaviour of the walls and ties and assessing this load share accurately is quite a complex task given the number of variables at play which included high ambient temperatures of 50°C with radiant effects giving rise temperatures of the order of 80°C in the exposed steel elements during construction.

Figure 7: Load path through structure

Whilst different threats to the integrity of the structure locally were carried out, including fire and the possibility of a terrorist act, we considered the scenario where each of the ties were lost. Such an extreme event would cause a large shift of load from the ties with the cores carrying all the thrust from the diagonals; it was therefore deemed prudent that both the ties and cores be designed to carry all the horizontal thrust from the diagonals thereby providing a structural system with 100% redundancy (Fig. 8). The matter of time-dependent load share was addressed by specifying a sequence of works which was developed in conjunction with Enerpac (a manufacturer of high-pressure hydraulic products - www.enerpac.com/en) in which the thrust into the concrete core was prevented by disengaging the strut into the concrete floor plate. This was achieved by boxing out the thick floor plate and staging the concrete pour locally in which a slotted splice connection in the anchor strut prevented load transfer to the cores. The tie beams, which centrally consist of a W36x800 (915mm deep x 1190kg/m) US steel sections, has a slot at mid-span and is connected by flange and web plates with longitudinal slots which allow the connection to move during pre-tensioning through strand-jacking across the slot (Fig. 9). This pre-tension is synchronised to ensure zero horizontal movement at the support thereby locking-in a working load of 24MN into the tie which is transferred through a full strength splice with high-strength friction-grip bolts designed for no slip at ultimate loads and design with slots allowing an anticipated movement of 31mm during the tensioning process. The second stage of work locks in the back strut thereby achieving the planned load path and level of redundancy in the structure.

http://www.enerpac.com/en

Figure 8: Resultant forces at connections of braces to core walls

Figure 9: Lower brace to core connection and detail of pre-tensioned beam

LATERAL LOAD RESISTING SYSTEM The two core walls are 1000mm thick at basement level and thin out to 400mm at level 67, they are tied internally by conventionally reinforced concrete floor plates and beams. The large rectangular openings over the height and width of the three external faces of the walls lead to these being classified as punched walls, which behave as a two-way sway frame with 1.8m deep 2.0m long columns but with all the advantages afforded by a wall or closed tube.

The lift cores are located within the two primary concrete core walls: these are 250mm thick over their full height and carry gravity loads only and whilst by virtue of their relative stiffness they see a small portion of the lateral loads, these were not taken into consideration and the primary outer concrete core walls were designed to resist 100% of the lateral loads.

The walls have large thickenings in each corner measuring 2.5m square and reducing progressively over height to 1.5m square and additionally on the vertical face the central row of mega-braces are connected to a central concrete column which at lower levels morphs into a 1400x1000 thickening in the core walls. The grade of concrete varies from C85 a basement to level 12, thereafter C75 to level 35, C65 to level 48 and 45 to level 65 all with a triple blend design mix similar to that for the foundations but with an upper limit of 70% GGBFS cement replacement.

The lateral stability for the building is provided solely by the two concrete cores acting in two different modes. In the transverse direction the cores are free standing vertical cantilevers each resisting 50% of the transverse loads with the floor plates distributing lateral loads to the core walls. The longitudinal releases in the columns described previously prevent the artificial stiffening of the upper portion of the core walls by the steel infill panels between the core walls. In the longitudinal direction stability is achieved through the pairing of the core walls by achieving a push / pull effect in the core walls with the steel diagonals and tie beams transferring the shear. The design of the mega-braces are governed by gravity loads with the transient load component arising from the worse of wind and seismic forces and which consisted of no more than 20% of the factored gravity loads.

The design evolved considerably for the initial schematic stage with the biggest challenge being the effect of large window openings which reduced considerably the stiffness in the core walls having an immediate impact on the building drift and accelerations. As with most super-tall buildings in moderate seismic zones the governing lateral loads were due to wind loads with design overturning moments of 9,530 and 3,860MNm in the transverse and longitudinal directions respectively but the base shear design was governed by seismic forces of the order of 59MN which was the trade off of the podium being connected to the tower (Fig. 10). As expected with a tall narrow structure the cross wind response effects are dominant with resonant effects consisting of 93%, 55% and 86 of the peak total moment. These dominant resonant effects were reduced with the stiffening of the upper steel lattice portion of the building supporting the viewing deck, in which fatigue was also a key design consideration.

Figure 10: Distribution of moment and shear forces over height of tower

The modal response of the structure is 6.21s (which lies within the expected band of H/46 to H/60) in the first mode and 3.50s and 3.18s in the second and third modes respectively, with the third mode being essentially a pseudo-torsional mode arising from the opposing first modes in the cores resulting in an overall torsional effect (Fig. 11). The key design criterion for this building was achieving the appropriate levels of occupant comfort which is discussed further.

Figure 11: First three fundamental modes: left to right with periods of 6.21s, 3.50s & 3.18s

DESIGN OF MEGA BRACES The mega-braces are fabricated steel box-sections measuring 600x660mm and 700x700mm made up of plate thicknesses varying from 50 to 85mm at the lower portion of the tower and reduce to the larger of the rolled H sections at the upper levels. The mega-braces are connected to the core walls through stiffened gusset plates up to 115mm thick site welded to a cast-in plate connected to large UC and W14x16 sections embedded horizontally and vertically into the concrete core walls. The embedded sections were sized to distribute stresses into the concrete cores through 25mm diameter headed shear studs arranged in rows of 10. The design of the studs were carried out with reference to EC4 (Eurocode 4)[14] which gives extensive guidance on the design of studs in higher strength concrete, this was then limited to 50% of the concrete capacity to ensure strain compatibility and maximise stud stiffness at transfer.

The analysis assumed the mega-braces to be continuous over their length with connections to the core walls assumed to be pinned with only the spandrel beams to column connections being rigid. All elements were arranged to ensure forces at all joints were concentric to avoid the possibility of secondary bending effects which due to the large axial forces at play would give rise to large bending moments for even small eccentricities. There was great internal debate about the assumption that the mega-truss behaves as a pin-jointed structure as the axial rigidity was far greater that the flexural rigidity of the primary truss elements. Based on the approach described by Smith and Coull[5] and a number of sensitivity checks carried out to assess the difference in stress levels validated the pin-jointed assumption.

FLOOR SYSTEM Typically in a tall building the two areas where an economical design achieves the most savings are in the core walls (or lateral load resisting system) and in the floor plates. These two elements often account for between 70 and 80% of the self weight of a tall building with any economy or over-conservatism amplified over a number of floors due to the repetitive nature of the design. This is particularly true for a building where 58 floors sit on a transfer structure spanning 40m.

The floor plate selected consists of on overall 160mm thick light weight concrete slab that acts compositely with the supporting steel beams which span between the primary structures formed by the three planes of mega-braces which span between the concrete core walls (Fig. 12). This floor plate is a cast in-situ concrete slab on an 80mm high profiled sheet deck reinforced with a steel mesh (or with a fibre reinforcement option to suit the contractor's preference) to achieve adequate durability and fire resistance. Additional reinforcement is required in the troughs of the sheet decking in more highly loaded areas to meet the NFPA requirements for a 2Hr fire resistance without supplementary sprayed fire protection to the soffit of the deck.

Figure 12: Typical floor layout

A floor plate system is capable of spanning up to 4.0 metres without propping during the construction stage was selected and a lightweight aggregate and a low concrete of strength (C25) concrete was specified which reduced the total self-weight of the floor plate by 20% and reduced the embodied carbon for a floor plate type which is repeated over 58 floors. Although more expensive than normal weight concrete, the benefit of the lightweight concrete option was the commensurate saving on the supporting steel structure which far outweighed the cost premium. In an additional effort to keep the self-weight of the building down all partitions and service shafts were specified to be light weight construction such as fire resistant dry wall and light weight blockwork with a density of less than 750kg/m3.

In addition to carrying gravity loads, the floor plates have additional functions: they act as a diaphragm spanning horizontally between the concrete core walls and carry small compression loads arising from the tilt in the facade, with these forces transferred though the

slab and balanced by virtue of the symmetry of the structural arrangement. The stresses in the floor plates were verified against the more critical effect of either wind load or when it acts as a collector element under seismic loading - the seismic loads governed the design at the lower floor levels with wind load being more critical at higher levels where the wind loads were higher and the floor plates narrower.

STEEL FLOOR BEAMS The initial floor beam option at the concept design stage was based on the principle of providing a clear services zone which meant that the aim of the structural design was to minimise the beam depths. With floor to floor heights of 4.0m and 150mm raised floor, 160mm slab, 50mm ceiling tiles and an allowance of 600mm for services, the longer floor spans of 13.5m would require 356x406x235UC sections which achieves a 2.590m floor to ceiling height which is less than the 2.7m minimum expected for grade A office space.

This separation of structure and services zones ended up with a floor structure that weighed over 60kg/m2 and prompted us to consider the alternative which meant a planned, flexible services reticulation scheme within the beam depth and with sufficient resilience to accommodate future changes and upgrades. This meant a high level of co-ordination and a thorough understanding of the different services being incorporated within the structural depth. We revisited our initial assumptions and considered different floor beam arrangements to come up with what we believe to be an optimum floor system through detailed analysis of cost and construction with data provided from a number of local steel fabricators. The various beam arrangements were compared against a baseline and were tabulated and indexed to allow for direct comparison of different options in terms of cost and tonnage.

The beam options considered were:

1. UB Sections (baseline for comparison) 2. UC Sections (the initial design) 3. Shallow trusses 4. Proprietary cellular beam plate girder system (Fabsec) 5. Proprietary cellular beam option from off the shelf profiles (Westok) 6. Fabricated plate girders (PG) 7. Deep UB section with openings as required

This assessment was carried out for a floor layout with beams centred at 2.67m and 4.0m and the results are summarised in Table 1 and Table 2.

This tabular format allowed for various assessments to be carried out including the embodied energy and environmental impact amongst others. The clear conclusion from the tables were that from both a cost and tonnage perspective a fabricated plate girder solution was optimal and in the time since this assessment was first made, Fabsec has setup a production line in Dubai with costs aligned with the fabricated plate girder option.

In conclusion, 650mm deep fabricated plate girders weighing between 66 and 124kg/m were specified for typical floors and increasing to 166 and 181kg/m at plant-room floor levels which achieved considerable savings when compared to the original scheme. These beams were specified with 400mm diameter and 400mm x 900mm oval openings to accommodate services and comfortably achieved a floor to ceiling height of 2.8m. FLOORS IN CORES, BASEMENTS AND PODIUM The floor plates within the cores are conventionally reinforced concrete flat slabs with down-stand beams for additional support for the longer spans and within plant-rooms. A mixture of precast and cast in-situ slabs were chosen for the podium floor plates with hollow core slabs ranging from 265mm on average to 400mm for longer spans. The first floor is designed to cater for heavier loads various landscaping and other features and are precast 1000mm deep double T-slabs spanning 14.5m and sitting on a cast in-situ RC beams.

MODELLING TECHNIQUES The building was modelled using Etabs (by CSI)[11] for the general strength and serviceability checks with long term effects on the cores assessed using MIDAS (by Midas Information Technology)[12]; both of these are very powerful software packages that carry out very complex analyses and provide extensive and very detailed output. These and other similar programs are important tools in the structural engineer's arsenal but have to be used with caution as they can lead to a false sense of security in the misplaced belief that they 'do it all' and often lead to a deferral of engineering judgement. A good understanding of engineering principles coupled with a 'feel' for structural behaviour is a key skill in establishing structural form and understanding the load path through more complex structures.

There are a number of modelling techniques that have been developed for the analysis of tall buildings that whilst applicable to more typical high-rise building designs, these need to be adopted with care in the design of super-tall or other more complex structures. The use rigid

Table 1 Table 2 Floor Beam Type Weight of

Steelwork in kg/m2 to Level 31 (22 No.

Floors)

Weight of Steelwork in kg/m2 to level 65 (36No.

Floors)

Average weight of Steelwork in kg/m2

Tonnage Index (Total)

Floor Beam Type Price Indexed to Level 31

Price Indexed to Level 67

Weighted Price Index (Total)

Beams at 4.0m c/c Beams at 4.0m c/cUB Beams (Baseline) 34.2 24.2 28.9 1.05 UB Beams (Baseline) 1.00 1.00 1.00UC Beams (356x368x129UC ) 46.6 24.2 34.6 1.26 UC Beams (356x368x129UC ) 1.96 1.38 1.40Trusses (Not Practical) 37.1 24.2 30.2 1.10 Trusses (Not Practical) 1.28 1.29 1.13Fabsec Beams 31.7 24.2 27.7 1.01 Fabsec Beams 1.48 1.74 1.25Westok Beams 31.7 24.2 27.7 1.01 Westok Beams 1.40 1.65 1.21Fabricated PG (****) 33.3 24.2 28.4 1.03 Fabricated PG (*) 1.30 1.45 1.15Deep UB Section 36.5 24.2 29.9 1.09 Deep UB Section (**) 1.26 1.79 1.17

Beams at 2.67m c/c Beams at 2.67m c/c (Simple Beams Above Level 32)Fabricated PG above level 30 26.3 28.2 27.5 1.00 Fabricated PG above level 30 1.42 1.17 1.26Deep UB Sections 49.1 28.2 36.1 1.31 Deep UB Sections 1.82 1.17 1.42UCs (Cambered) 55.5 28.2 38.6 1.40 UCs (Cambered) 1.79 1.17 1.41

UBs (Cambered) 36.3 28.2 31.3 1.14 UBs (Cambered) (***) 1.15 1.17 1.16

Tonnage Comparisons

Of the Three in Bold(****) - Lowest Tonnage PG - Plate Girder UB - Universal Beam UC - Universal Column

Indexed Price Comparison

Of the Three in Bold(*) - Lowest Comparative Costs - Higher Embodied Energy and CO2 emissions(**) - Highest Comparative Costs - Lowest Embodied Energy and CO2 emissions(***) - Highest Tonnage of steelwork - Best Balance of Cost and Environmental Impact (Note: Lowest Fabrication Costs and Embodied Energy, But High Tonnage = High Energy Expenditure For Erection)

diaphragms (master/slave commands) that simplify the analysis and shorten the running time of a model and effects of switch on gravity need to be understood and investigated thoroughly as these are likely to skew the results of complex transfer structures carrying large gravity loads.

A single model approach is inappropriate when assessing the effects of a large number of variables such as loading history (which tends to be done sequentially floor by floor and carried out towards the end of the design process) and the sensitivity to the stiffness properties assumed for the cores and which governed the load distribution between the core walls and the tie beams. Multiple models we created in an upper and lower bound approach to assess the forces in the steel sections for cracked and un-cracked concrete properties with the true behaviour lying somewhere between the two extremes.

The loading history was accounted for by considering the steel columns within the diamonds as being pinned to and longitudinally released from the underside of the mega-braces. As described earlier, this also had the benefit of overcoming the effect of switch-on gravity, a phenomenon in modelling where the loading and stiffness history of a structure are assumed to be synchronised and instant. The load paths were checked by applying a set of dummy loads of 10MN at key parts of the structure at different levels; this allowed us to better understand how loads moved through the structure and allowed us to gauge what we expected intuitively against what was being observed. This also confirmed that our design assumptions and understanding of the behaviour were correct.

When modelling tall buildings the wind loads obtained from the wind tunnel report are usually applied at the centre of stiffness of the structure which in this case was at the centroid of the office floor, half way between the cores. The floor plates were included in the model to correctly account for the load and lateral stiffness of the floor plate but the stiffness was modified in the longitudinal direction so that the correct assessment of load and stiffness of the in the steel elements could be made. The wind loads were divided in two and applied at the centroid of the cores to removed parasitic forces in the steel frame with torsional effects taken as a complementary push and pull taken as torsional moment divided by lever arm.

The stresses in the mega-braces resulting from the shortening of the cores between successive connection points were included for in the axial force outputs but these accounted only for the elastic shortening component. Time dependent effects due to shrinkage and creep were assessed in accordance with the approach by Fintel, Ghosh and Iyengar[8] resulting in shortening of approximately 150% the elastic value and a stress range of 14 to 85 N/mm2 in the mega-braces. These values were in agreement, although slightly conservative, when compared to the results obtained in the MIDAS model.

WIND ENGINEERING AND OCCUPANT COMFORT Wind induced loads and occupant comfort were at the forefront of all considerations throughout the design process. CPP was the wind engineering consultant who used a high frequency base balance model to establish the static and dynamic wind loads on the building and to assess occupant comfort (Fig 13). An early concern was the comfort of the patrons at viewing deck level and the response due to modal decoupling of the lattice tower from the concrete cores below.

The global wind effects on the building were described earlier and did represent the governing design loads on the building. This translated into peak cladding pressures of between -2.5 to +2.5kPa with peak pressures of 3.5kPa. Peak accelerations at the uppermost

occupied floor was 16-20mg depending on the wind direction for a 1 in 20 year return period range of damping values of 1.0 to 2.0%, which is within the acceptable range of up to 21mg for office type buildings. This was checked against the criteria outlined in ISO 6897[15] and NBCC[16] for different wind return periods of 5 and 10 years respectively and were again found to be within acceptable limits.

The viewing deck level on the other was found to have accelerations of between 30 and 40mg which was deemed acceptable as the deck would not be accessible during high winds. This was achieved by limiting access to the deck at times where wind speeds exceed 22m/s by controlling the entrance through doors and the panoramic lift electronically through a unit linked to an anemometer.

Figure 13: Test model in wind tunnel with ground roughness and adjacent buildings

simulated (photo courtesy of CPP) COMPUTED DRIFT AND DRIFT LIMITS The building is designed to US standards (ASCE07/05, ACI318/05, AISC 341 and 360/05 and UBC97)[17/18/19/20/21] which set drift limits under wind loads to between H/400 to H/600. Later revisions of these codes relax these limits to H/400 which is not consistent with inter-storey drift limits of H/500. The difficulty in using international codes is that they tend to be interpreted quite narrowly particularly when the context of the country of origin is lost on the country adopting the code. When designing a tall building which lies outside code parameters it is always advisable to engage with both the peer reviewer and local authority at a very early stage in order to identify critical items and indicate how these will be overcome.

Meeting the overall and inter-storey criteria under a strict reading of the codes was an issue from the outset and a practical approach was required. The local authority was in agreement that satisfying occupant comfort was the key criterion provided the integrity of the cladding and partitions were maintained. Under the code of combination of D + 0.5L + 0.7W stipulated in ASCE 07/05 the total drift at the crown of the tower was 1055mm which equates to an overall drift of H/381 which was not deemed compliant. However, by splitting the building into two components with the drift assessed at level 67 as H/581 and the drift of the lattice tower at H/215 complying with BS8100[22], satisfied the authorities.

Code stipulated storey drifts are stringent, even more so when taken at face value from analysis software which computes drift ratio as the horizontal component of the drift over height. This notion when applied to tall building leads to gross over-conservatism as it tends

to find the maximum inter-storey drift in the upper portion which is counter intuitive. When considering the effect of rigid body rotation [8/9/10] - shortening of the compression fibre and lengthening of the tension fibre - which takes into account that the floor are not level leads to the accurate measure of the racking shear.

DESIGN OF CONNECTIONS The steel to concrete connections were designed from first principles making use of hand calculations to establish the load path from source to point of resistance. The critical connections of mega-braces to concrete core walls were modelled using SAP2000 (by CSI)[13] which with use of in-plane stiffness modifiers to model a reliable load path through the floor plates into the core walls (Fig. 14). The level 8 floor slab consists of an 800mm thick RC floor plate design to provide 100% redundancy by transferring the horizontal component of the loads in the lower central mega-brace into the core walls. The floor slab was reinforced with a combination of T40 diagonals to transfer the shear and an embedded W14x16x605 H section to transfer the tension forces into the concrete cores.

Figure 14: Typical mega-truss to concrete core connections and steel node connections

KEY ELEMENTS AND PROGRESSIVE COLLAPSE The large transfer structure which spans 40m between concrete cores can accommodate upwards of 5000 people and does require that the effects of disproportionate collapse be considered in some detail. We followed guidance outlined in the England and Wales Building Regulations - Document A3[23] and in BS 5950[24], which is quite informative and thorough. From this it was concluded that there was sufficient redundancy and alternative load paths for none of the structural member to be classified as key elements. We demonstrated this to the approving authority by running an analysis with one portion of the diagonals between floor plates in each of the three planes of the mega-trusses removed. This confirmed the high level of redundancy within the structure with the remaining members having sufficient capacity and ductility to carry this additional load.

CONCLUSION This paper presented the key design features behind what is a very innovative structure set to be located at the heart of the Financial District in Dubai. It gives an overview of the thought process behind key decisions taken during the design, leading to what is a relatively simple and buildable solution. ACKNOWLEDGEMENTS The author would like to acknowledge the contributions of Bob Scott, Ranjith Chandunni, Min-Hah Chun, Martin Holford, Farhad Pazoki, Karsten Veith, Ahmed Adb Elazziz Hussein and Farhad Homayoun Shad who were amongst many within the Atkins team to have contributed to this project at the various stages of the design. The author would also like to thank the client, DIFC, for their support in the writing and publishing of this paper.

REFERENCES [1] Architecture and Urban Design - Selected and Current Works 2010, Atkins.

www.atkinsglobal.com www.middle-east.atkinsglobal.com [2] Berahman, F. and Esther, M. J. “Hybrid Solution for tall building structures” presented at

CTBUH 2010 Mumbai Conference, available online at: http://www.ctbuh.org/Events/Conferences/Mumbai2010/Presentations/Mumbai2010T20SpeakersPresentations/tabid/1459/language/en-US/Default.aspx

[3] Berahman, F., Esther, M., Scott, R., Chandunni, R. and Veith, K., “The Structural Design of Icon Hotel in Dubai, United Arab Emirates” The Structural Engineer, Proceedings of the Institution of Structural Engineers, 89(7), 26-32, 2011.

[4] Berahman, F. “Performance based seismic design of Icon Hotel in Dubai” presented at LATBSDC annual conference in Los Angeles on 7th of May 2010 and published online in Wiley online library in the special issue of Journal of Structural Design of Tall and Special Buildings, December 2010. http://peer.berkeley.edu/tbi/events.html.

[5] Stafford Smith, B. and Coull, A. “Tall Buildings Structures: Analysis and Design,” John Willey and Sons, Inc. 1991.

[6] Naeim, F. “The Seismic Design Handbook,” Kluwer Academic Publisher, USA, 2011. [7] Fintel, M., Ghosh, S.K. and Iyengar, H. “Column Shortening In Tall Building Structures-

Prediction and Comparison,” Portland Cement Association, 1987. [8] Beedle, L. "Structural Systems for Tall Buildings,” International Edition, McGraw-Hill

1995. [9] Seismic Design Guide "Recommendations for the Seismic Design of High-Rise

Buildings", Council for Tall Buildings and Urban Habitat 2008. [10] Mendis, P., Ngo T., Haritos N., Hira A., Samali B. and Cheung J., “Wind Loading on

Tall Buildings”, EJSE Special Issue: Loading on Structures.1-14, 2007. [11] Computers and Structures Inc (CSI) “ETABS nonlinear version 9.5, extended 3D

analysis of building systems” CA, USA, 2008. [12] MIDAS Information technology “MIDAS Gen Version 7.4: User manual” 2008. [13] Computers and Structures Inc (CSI) “SAP 2000 Linear and Nonlinear Static and

Dynamic Analysis and Design of Three-Dimensional Structures V11” CA, USA, 2006.

http://www.atkinsglobal.com/http://www.middle-east.atkinsglobal.com/http://www.ctbuh.org/Events/Conferences/Mumbai2010/Presentations/Mumbai2010T20SpeakersPresentations/tabid/1459/language/en-US/Default.aspxhttp://www.ctbuh.org/Events/Conferences/Mumbai2010/Presentations/Mumbai2010T20SpeakersPresentations/tabid/1459/language/en-US/Default.aspxhttp://peer.berkeley.edu/tbi/events.html

[14] BS EN 1994-1-1:2004 " Eurocode 4: Design of Composite Steel and Concrete Structures - Part 1-1: General Rules and Rules for Buildings" BSI 2004.

[15] ISO 6897:1984 "Guidelines for the evaluation of the response of occupants of fixed structures, especially buildings and off-shore structures, to low-frequency horizontal motion (0,063 to 1 Hz)".

[16] NBCC, "The National Building Code of Canada", The Institute for Research in Construction (IRC) 2005.

[17] ASCE7-05, "Minimum Design Loads for Buildings and Other Structures," American Society of Civil Engineers, 2006.

[18] ACI318/05, "Building Code Requirements for Structural Concrete and Commentary," An ACI Standard, The American Concrete Institute, 2005.

[19] ANSI/AISC 341/05, "Seismic Provision for Structural Steel Buildings," American Institute of Steel Construction, 2005.

[20] ANSI/AISC 360/05, "Specification for Structural Steel Buildings," American Institute of Steel Construction, 2005.

[21] International Conference of Building Officials (ICBO), “Uniform Building Code” Whittier, 1997, CA.

[22] BS 8100-1:1986 "Lattice Towers and Masts - Part 1" BSI 1986. [23] "The Building Regulations 2000 - Document A3" 2004 Edition, NBS 2006. [24] BS 5950-1:2000 "Structural Use of Steelwork in Building - Part 1: Code of Practice for

Design - Rolled and Welded Sections", BSI 2000. [25] Robertson, L., "Gold Medal Address, Presented at The East Winter Garden, Canary

Wharf, 9 March 2005" The Structural Engineer, Proceedings of the Institution of Structural Engineers, 20-27, 2005.

ABOUT THE AUTHOR

Matthew J. Esther BEng (Hons) CEng EurIng FIStructE MICE MSFE

Associate and Senior Structural Engineer WS Atkins and Partners Overseas, PO Box 5620 Dubai UAE matthewesther@hotmail.com / matthew.esther@atkinsglobal.com

Matthew J. Esther is a Chartered Civil and Structural Engineer whose areas of expertise are in the design of tall buildings, long span structures, geometrically complex roof structures and stadia. He has previously worked in the UK and Greece and is currently based in Dubai, UAE where he is an Associate with WS Atkins.

Matthew has held senior design positions for a number of years where he has been responsible for the technical direction, management and delivery of a range of large projects. His key project experience to date also include Trump International Hotel and Tower, Icon Hotel, Anara Tower, Gulf News HQ (UAE), ITC Tower and Namaste Tower (India), Padideh Shandiz Waterpark and the Esfahan Convention Centre (Iran), Lusail LRT and Doha Metro (Qatar), the Hellinikon Olympic Stadium and the Thessalonica Metro (Greece), the Millennium Dome, Wembley Stadium, the Hertfordshire University Campus, Vertex Chineham Court and Thames Barrier Point (UK).

Matthew has been an active member of the UAE Division of the Institution of Structural Engineers since its inception and has served in various capacities as committee official which included tenure as Chairman in 2011. He is the organiser and principal lecturer for the Chartered Membership Exam Preparation Course in the UAE and in 2011 was the first recipient of the President’s Award for contributions to the advancement of the Institution in the region. He lectures frequently at Heriot-Watt University in Dubai, on their BEng / MEng Civil Engineering course.