Polastri A. 1 PROPOSAL FOR A STANDARDIZED DESIGN AND ...

International Journal for Quality Research 10(3) 607–624

ISSN 1800-6450

607

Polastri A. 1

Pozza L.

Article info:

Received 17.02.2016

Accepted 18.08.2016

UDC – 54.061

DOI – 10.18421/IJQR10.03-12

PROPOSAL FOR A STANDARDIZED DESIGN

AND MODELING PROCEDURE OF TALL

CLT BUILDINGS

Abstract: A crucial issue in the design of a mid-rise Cross

Laminated Timber (CLT) building under horizontal seismic

action, is the definition of the principal elastic vibration period

of an entire superstructure. Such vibration period depends on

the mass distribution and on the global stiffness of the

buildings. In a CLT structure the global stiffness of the

buildings is highly sensitive to deformability of the connection

elements. Consequently for a precise control of the vibration

period of the building it is crucial to define the stiffness of each

connections used to assemble a superstructure. A design

procedure suitable for a reliable definition of the connection

stiffness is proposed referring to code provisions and

experimental tests. Discussion addresses primary issues

associated with the usage of proposed procedure for numerical

modeling of case study tall CLT buildings is reported.

Keywords: CLT structures, core structures, seismic design,

shear walls, tall buildings

1. Introduction1

In recent years the Cross Laminated Timber

(CLT) panels have become widely employed

in Europe and elsewhere to build multi-story

residential and commercial buildings. These

buildings are often characterized by the

presence of many internal and perimeter

shear walls. Such typology has been widely

studied through experimental tests and

numerical simulations methods.

The most comprehensive experimental

investigation to date on seismic behaviour of

CLT buildings was carried out by CNR–

IVALSA, Italy, within the SOFIE Project

(Ceccotti, 2008; Ceccotti et al., 2013). Other

important investigations have been

1 Corresponding author: Polastri Andrea

email: [email protected]

conducted at the University of Trento, Italy

regarding CLT structures (Tomasi and

Smith, 2015) and hybrid steel-CLT

technologies (Loss et al., 2015). European

seismic performance related tests have also

been conducted at the University of

Ljubljana, Slovenia where the behaviour of

2-D CLT shear walls with various load and

boundary conditions were assessed (Dujic et

al., 2005). FPInnovations in Canada has

undertaken tests to determine the structural

properties and seismic resistance of CLT

shear walls and small-scale 3-D structures

(Popovski et al., 2014). Those and other

studies have enabled characterization failure

mechanisms in large shear wall systems

(Pozza and Scotta, 2014).

Mostly traditional timber structural systems

have employed post and beam arrangement

to resist effects of gravity loads, while

effects of lateral loads are resisted by cross-

mailto:[email protected]

608 A. Polastri, L. Pozza

bracing or non-timber frame infill material

(Smith et al., 2009; Tlustochowicz et al.,

2010; Bath, 2013). In the modern context an

equivalent construction technology is to use

beams and columns to resist effects of

gravity loads, while CLT building core

substructures and other CLT shear walls to

resist effects of lateral loads (i.e. earthquake

or wind). Such structures are structurally

effective, and fail in predictable stable ways

if overloaded (Smith et al., 2009).

Advantages of such systems include the

creation of large open interior spaces, high

structural efficiency, and material saving.

Although examples of the new typology

have been built already, there has not been

full study of the structural behaviour. This

means that there is not yet clear

understanding of optimal structural

configurations, dimensional limitations of

spans, minimum number of columns or

maximum number of storeys that is feasible.

A preliminary investigation about the

response of such building typology is firstly

reported in (Polastri et al., 2014) where the

behaviour of multi-storey buildings braced

with CLT cores and additional shear walls

was examined based on numerical analyses

of various 3-dimensional configurations

adopting two different calibration of the

numerical model according to codes and

experimental test data respectively

Researcher investigate the seismic behaviour

of tall CLT building developing new

technologies and hybrid steel-timber

structures (Ashtari et al., 2014; Bath et al.,

2014; Liul et al., 2014). However the most

crucial aspects relate to the construction

method for CLT building cores and the

pertinent issues relate to vertical continuity

between storeys, connections between

building core elements and elevated floors,

and core to foundation connections is not

completely investigated yet.

Some additional studies are reported in

Polastri et al. (2015). In this work a proposal

for a standardized procedure to realize a

reliable design of CLT superstructures was

presented and validated by means of modal

response spectrum analyses on various

building configurations. In addition the issue

of diaphragm in plane behaviour is treated

and the interaction with CLT wall detailed.

Despite several studies, it has to be noted

that available seismic codes do not provide

guidance on the most crucial aspects of how

to design structural systems that combine

post and beam type frameworks with CLT

building cores and shear walls (Smith et al.,

2009). The primary issue is that, the

estimation of the principal vibration periods,

of buildings with Seismic Force Resisting

Systems (SFRS) containing CLT wall

panels, can be grossly inaccurate if proper

attention is not paid to accurate

representation of connection stiffnesses.

Estimates of principal elastic period (T1)

obtained using the simple formula in

Eurocode 8 (CEN, 2013) can deviate greatly

from values found using finite element

models employing connection stiffnesses test

data. Similarly finite element model

predictions of T1 in which connection

stiffnesses are estimated from information in

Eurocode 5 (CEN, 2014) can differ greatly

from values obtained using connection test

data. Inaccurate representation of connection

stiffnesses can also result in incorrect sizing

of elements in SFRS, and gross inaccurate in

predictions of inter-story drift.

The aim of this work is provide design

guidance related to determination reliable

principal elastic vibration period of CLT

superstructures starting form proper

definition of initial stiffnesses as well as

capacities of connections. Therefore a

suitable calculation process for design of

CLT SFRS is developed, starting form

procedure detailed in Polastri et al. (2015).

2. Design and modelling procedure

for tall CLT buildings

A crucial issue in the design of a CLT

building under horizontal seismic action, is

the definition of the principal elastic

609

vibration period (T1) of an entire

superstructure (CEN, 2013, Lucisano et al.,

2016). Such vibration period depends on the

mass distribution and on the global stiffness

of the buildings. In a CLT structure the

global stiffness of the buildings is highly

sensitive to deformability of the connection

elements (Pozza et al., 2015). Consequently

for a precise control of the vibration period

of the building it is crucial to define the

stiffness of each connections used to

assemble the superstructure. During the

design process engineers are required to find

following a iterative procedure the principal

natural frequency (f1 = 1/T1). Figure 1

represents the proposed iterative procedure:

(1) the stiffness of the connections

influences the global stiffness of the building

and therefore its principal elastic period; (2)

the external force induced by earthquake in

each connection is a function of the principal

vibration period; (3) the uniaxial load

bearing capacity of the connection must be

compatible with the external force; (4) the

strength and the stiffness of the connection

are linked through the effective number of

fasteners. In addition: (5) the interaction

between the tensile and shear force (or

displacement) acting on the connection must

be verified in order to avoid working

condition inconsistent with the resistant

domain of the connection. This interaction

can be checked adopting the resistant

domain prescribed by connection

manufactory (e.g. EOTA) or referring to

specific experimental tests that investigate

the bi-directional behaviour of the

connections. An experimental campaign

aimed to define such interaction is on-going

at CIRI Buildings & Construction

Laboratory of the University of Bologna –

Italy.

Figure 1. Calculation process for design of connections

According to the scheme in Figure 2, an

efficient approach to design a CLT structure

is to start from a preliminary definition of

the external force induced by earthquake in

each wall panel according to the common

equivalent static force linear static analysis

approach (CEN, 2013, Pavlovic and

Fragassa, 2016).

Such definition follows the engineering

design practice disregarding the effect of the

interaction between tensile and shear action

on the connection resistance. A more

accurate approach may consider the effective

capacity of connection in both tensile and

shear. However no experimental results are

available to define the interaction between

tensile and shear connection capacity

therefore such approach is applicable only

for research purposes (Pozza et al., 2015).

The preliminary calculation does not involve

the definition of T1 accounting for effects of

connection stiffness but refers to a priori

definition (e.g. according to CEN, 2013) of

the periods referring just to the number of

storey and to the building typology.

Once static forces on each CLT wall panel

are defined connection capacities can be

designed to be compatible with external

static forces. This allows estimation of the

connection elastic stiffness and therefore

realistic preliminary estimation of T1 using a

more precise Natural Frequency analyses in

a specifically developed building numerical

model. Then T1 can be used in modal

CONNECTIONS ELASTIC

STIFFNESS

BUILDING PRINCIPAL

ELASTIC PERIOD

EARTHQUAKE FORCE ON

CONNECTIONELEMENTS UNI-AXIAL

CONNECTION LOAD CAPACITY (check)

INTERACTION BETWEEN CONNECTION TENSILE AND

SHEAR FORCE (check)


analyses to calculate the effective forces

induced in connections by earthquakes.

Obtained connection forces may or may not

be compatible with the connection strength,

and if not it is necessary to redesign

connections. Finally it is necessary to verify

the compatibility of the bi-directional action

and displacement of the connection with the

relative interaction domain. In detail the

displacement achieved by connection should

be verified bot for tensile and shear actions

in order to avoid brittle failures or relevant

strength reductions.

Figure 2. Calculation process for design of connections

Afterward it is possible to perform a more

iterative precise frequency analyses until

solutions, including connection designs.

Results obtained in different interaction

(until convergence) should be significantly

different as reported in section 4.5.

Proposed procedure is based on simplified

Linear Dynamic Analyses (i.e. Modal

Response Analyses) adopted to calculate

connection’s actions and displacements.

Finally it is necessary to highlight that when

more refined numerical models are used to

reproduce connection behaviour (e.g. those

used in Pozza and Scotta, 2014) and

Nonlinear Analyses are performed, some

steps of the proposed procedure are

unnecessary, in particular those regarding

the check of consistency of calculated action

PRELIMINARY DESIGN VIA LINEAR STATIC ANALYSESelastic period T1 and horizontal force distribution according to EC8 formulation

DEFINITION OF 1st (ith) ATTEMPT CONNECTION DISTRIBUTION

BUILDING MODELING ADOPTING THE CONNECTION ELASTIC STIFFNESS (kser - EC5 formula or ktest -

experimental test ) RELATED TO ith DISTRIBUTION

NATURAL FREQUENCY ANALYSES TO DEFINE THE ACTUAL VIBRATION MODE OF THE STRUCTURE

MODAL ANALYSES TO DEFINE THE ACTUAL FORCE ACTING IN EACH CONNECTION ELEMENTS

CALCULATED FORCE > (greater than)CONNECTION

STRENGTH

UPDATE CONNECTION

DISTRIBUTION (i+1)

[connection elements are re-distributed

among the building walls according to new force pattern]

CALCULATED FORCE < (lower than)CONNECTION

STRENGTH

CHECK CONSISTENCY BETWEEN RESISTANT

DOMAIN AND TENSILE/SHEAR INTERACTION

STOP

611

and one- or bi-directional capacity of

connections.

3. Connectors mechanical

characterization

The most important step in the procedure for

the design of CLT buildings, consist in the

reliable definition of the elastic stiffness of

the connections.

The connection’s elastic stiffness and

capacity can be evaluated referring to results

from monotonic or cyclic tests or to

analytical calculations according to

appropriate design code.

Below stiffness and capacity values

implemented into the numerical models

described in Section 4 were calculated

directly from experimental data and from

analytical formula provided by code.

3.1. Experimental characterization of

CLT metal connector elements

The mechanical behaviour of connection

systems for CLT structures that employ thin

metal elements fastened to panels with nails

or other slender metal fasteners is well

known, as demonstrated by numerous

studies conducted in the last years. The first

considered study was carried out at CNR-

IVALSA (Gavric et al., 2011), the second

study at the University of Trento (Tomasi

and Smith, 2015). Both tests were conducted

according to the European standard EN

12512 (CEN, 2006). The CEN 2006 protocol

provides a load history characterized by load

cycles of increasing intensity and is intended

to apply to structures in seismic regions.

Figure 3 reports the geometrical

characterization the fastening systems

adopted in the four examined connections:

angle brackets BMF 100 x 100 and

Rothoblaas TITAN TTF200 (EOTA, 2012)

used as shear elements and hold downs

Rothoblaas WHT 540 and WHT 620

(EOTA, 2011).

Connector angle brackets

BMF 100 x 100 angle TITAN TTF200

holdown WHT

540 holdown WHT 620

Geometric

al detail

n° and

type of

nails

n°12 Anker

4x60 n°30 Anker 4 x 60

n°12 Anker 4

x 60 n°32 Anker 4 x 60

Figure 3. Geometrical detail and fastener systems of investigated connections

Figure 4 reports the experimental load

displacement curve for holdown WHT 540

and angle brackets BMF 100 obtained by

Gavric et al. (2011). Figure 5 shows the

experimental load displacement curve for

holdown WHT 620 and angle brackets

TITAN TTF200 obtained by Tomasi et al.,

(2015) and reported in Polastri et al. (2015).

612 A. Polastri, P. Luca

a)

b)

Figure 4. Typical tests results: holdown WHT 540 (a) and angle bracket BMF 100 (b)

a)

b)

Figure 5. Typical tests results: holdown WHT 620 (a) and angle bracket TTF200 (b)

Since this paper deals with Linear Dynamics

Analysis of superstructure systems only the

parameters that characterize the maximum

load at failure (Fmax) and the elastic

properties of connections (ktest) are reported

here, Table 1.

The maximum load at failure is evaluated as

the peak force achieved during the tests

(CEN, 2006) and listed in Table 2 for the

examined connection elements.

The initial stiffness was firstly calculated

according to ‘method b’ specified by EN

12512 (CEN, 2006) that permits description

of the mechanical behaviour of trends

representing elastic phase and post-elastic

phase responses (Piazza et al., 2011). Such

approach define the elastic branch of bilinear

approximation of experimental curve

referring to the gradient between the point on

the load-slip curve corresponding to 0.1 Fmax

and the point on the load-slip curve

corresponding the 0.4 Fmax.

Since the elastic branch provided by

“method b” is conventional in some

particular cases it does not fit properly the

experimental elastic stiffness of the

connection. Consequently an additional

estimation of the experimental elastic

613

stiffens is provided referring to the actual

initial stiffens according to the results

provided by “method a” (CEN, 2006). The

experimental stiffness estimation of

examined connections are reported in Table

2 and in Figure 4 and Figure 5: dashed line

represents the stiffness calculated according

“method a” whereas continuous line

represents the stiffness according “method

b”.

3.2. Analytical definition of strength and

stiffness according to Eurocode 5

The lateral load bearing capacity of

connectors can be calculated according to

Johansen's yield theory (Johansen, 1949).

Standards (CEN, 2014) base formulas for

estimating the lateral capacity of slender

fasteners like nails on the aforementioned

theory. Here, the Eurocode 5 (CEN, 2014)

definition is used to calculate the

characteristic shear capacity per nail, Fv,Rk

(CEN, 2014). For the pertinent case, thick

steel plate, representing the vertical leg of

hold-down anchor or of angle bracket, the

load capacity per nail is (Equation 1).

c)

(1) d)

e)

Where: fh,k = characteristic embedment

strength in the timber member; t1 = depth of

penetration of the fastener into the timber

member; d = diameter of fastener; My,Rk =

characteristic fastener yield moment; Fax,Rk =

characteristic withdrawal capacity of the

fastener (associated with pulling the fastener

out of the timber member). Characteristic

embedment strength fh,k was computed

according to Eurocode 5 for nails without

predrilled holes (CEN, 2014) (equation 2):

fh,k = 0.082 d-0.3 k (2)

Where ρk is the characteristic value of panel

density. Fastener yield moment My,Rk and

withdrawal capacity of fastener Fax,Rk are

parameters computed according to the

manufacturers' technical certifications and

are therefore product-specific. Application of

equations (1) and (2) results in calculation of

a basic design resistance, which does not

include the necessary adjustments to obtain

the proper design resistances, taking into

account also the partial coefficients of

materials.. The design capacity per

fastener/nail is therefore computed according

to equation (3):

Fd, nail = Fv,Rk kmod / m (3)

Where kmod is the modification factor with

regards to the combined influences of

duration of loading and moisture; m is the

partial coefficient of the materials.

The analytical fastener stiffness is calculated

according to the formula given by Eurocode

5 (CEN, 2014) for steel to timber

connections reported in Equation 4.

Kser = 2 ∙ (m1.5 d0.8 /30) (4)

where ρm is the mean density of the panel

and d is the nail diameter.

The following values have been assumed

according to information on the material

property of fastener and CLT: My,Rk = 6.55

Nm; Fax,Rk = 1.32 kN; t1 = 55.6 mm; ρk =

350 kg/m3, ρm = 420 kg/m3. Moreover the

following design coefficients has been used:

kmod = 1.10, γm = 1.00, matching values

suggested by Eurocode 5 (CEN, 2014) in a

seismic design perspective. This results in

the predicted design values for lateral load

resistance Fd, nail, listed in Table 1 for the


three failure mode provided by Johansen's yield theory.

Table 1. Eurocode 5 derived nail capacity and stiffness

SINGLE NAILS (ANKER 4 x 60) mode c mode d mode e

Characteristic capacity Fv,k [kN] 4.06 2.16 1.93

Design capacity Fv,d [kN] 4.47 2.38 2.12

Slip moduli kser [kN/mm] 1.32

Starting from the load capacity and the

elastic stiffness of the nail it is possible to

calculate the analytical global properties of

the examined connections. In this case, due

to the large spacing of the nails, no reduction

effects, both for capacity and stiffness, are

considered.

In addition the initial stiffness of connectors

was calculated taking into account only the

stiffness of the steel-to-timber nailed joints.

The deformation of steel parts within the

connections has been neglected because it is

very small compared deformation of nailed

joints. Characteristic load-carrying

capacities, Fv,Rk, and slip moduli, kser are

listed in Table 2.

Table 2. Experimental and Eurocode 5 derived connection properties

CONNECTION

TYPE

Elastic stiffness (kN/mm) Capacity (kN)

Test (ktest) EC5 (kser) Test (Fmax) EC5 (Fy,Rd)

“method a” “method b”

BMF 100 3.2 2.5 7.3 23.5 14.3

TTF 200 9.8 8.2 23.1 70.1 35.5

WHT 540 6.1 4.5 15.8 48.3 25.4

WHT 620 16.3 12.1 42.2 100.1 67.8

The axial resistance and slip moduli

calculated above refer to the hypothesis that

the load carrying capacity of the nails was

weaker than the steel plate one according to

the capacity design principles in timber

structures (Fragiacomo et al., 2011).

Moreover the calculations follows the

engineering design practice disregarding the

effect of the lateral action on the holdown

resistance and stiffens. The analytical

procedures provided by codes don’t account

for the elastic and plastic deformability of

the steel plates and base bolts consequently

the calculated stiffness values result higher

than those obtained from experimental tests.

4. Numerical analysis of core tall

buildings

The behaviour of multi-storey buildings

braced with cores and CLT shear walls is

examined using numerical modal response

spectrum analyses, with connection

properties calibrated based Eurocode 5

(CEN, 2014) and experimental test discussed

in Section 3. Analyses followed the scheme

in Figure 2 and are presented in terms of

principal elastic periods, base shear and up-

lift forces, and inter-storey drift. In addition

the variability of these results will be

presented in the various interactions of the

methods, still the convergence.

4.1. Case study buildings

The aim is to characterize behaviour of

multi-storey CLT buildings braced with

cores and additional shear walls from the

seismic design perspective based on effects

of varying design parameters. Varied design

parameters are: number of storey (3-5-8),

lateral shear wall panels width (i.e. jointed or

un-jointed wall panels), construction

methodology (i.e. storey by storey shear-

walls or multistorey shear-walls), and

regularity of connectors as a function of the

615

height within a superstructure, Table 3.

Table 3. Examined building configuration

Case study

ID

3(5-8) A R 3(5-8) A I 3(5-8) B R 3(5-8) B I 3(5-8) C R

Graphical

schematizati

on of

building

cores (ex. 3-

storey case)

Panel

assembly Unjointed wall panels Jointed wall panels

Unjointed

wall panels

Elevation

regularity Regular Irregular Regular Irregular Regular

Constructio

n

methodolog

y

Storey by storey shear wall system

Multy-storey

shear wall

system

4.2. Geometric configurations

Examined case-study building

superstructures have footprint dimensions of

17.1m (direction X) by 15.5m (direction Y).

Seismic Force Resistant Systems (SFRS)

include a building core that is 5.5m by 5.5m

on plan, and partial perimeter shear walls

constructed from CLT panels with a total

base length of 6m. Storey height is 3m in all

cases, resulting in total superstructure

heights of 9m, 15m and 24m respectively (3-

5-8 configurations). All CLT panels in the

core walls have a thickness of 200mm. CLT

panels in perimeter shear walls are 154mm

thick, except for those in the lowest four

storeys of the 8-storey SFRS which are

170mm thick. Floor diaphragms are

composed of 154mm CLT panels in all

cases. Finally the distribution of connection

elements adopted among the height of the

examined building configurations is reported

in Table 4.

Table 4. In height connections distribution for the examined building configuration BUILDING

ID

CONNECTION

TYPE

CONNECTION LINE AMONG THE BUILDING HEIGHT

Base Floor 1 Floor 2 Floor 3 Floor 4 Floor 5 Floor 6 Floor 7

3 A R HOLDOWN

ANGLE BRACKET

WHT

620 TTF

200

WHT

540 BMF

100

WHT

540 BMF

100

- - - - -

3 A I HOLDOWN ANGLE

BRACKET

WHT 620

TTF

200

WHT 540

BMF

100

WHT 540

BMF

100

- - - - -

3 B R HOLDOWN

ANGLE

BRACKET

WHT

620

TTF 200

WHT

540

BMF 100

WHT

540

BMF 100

- - - - -

3 B I HOLDOWN

ANGLE

BRACKET

WHT

620

TTF

WHT

540

BMF

WHT

540

BMF

- - - - -


200 100 100

3 C R HOLDOWN

ANGLE BRACKET

WHT

620 TTF

200

- - - - - - -

5 A R HOLDOWN

ANGLE BRACKET

WHT

620 TTF

200

WHT

620 TTF

200

WHT

540 BMF

100

WHT

540 BMF

100

WHT

540 BMF

100

- - -

5 A I HOLDOWN ANGLE

BRACKET

WHT 620

TTF

200

WHT 620

TTF

200

WHT 540

BMF

100

WHT 540

BMF

100

WHT 540

BMF

100

- - -

5 B R HOLDOWN ANGLE

BRACKET

WHT 620

TTF

200

WHT 620

TTF

200

WHT 540

BMF

100

WHT 540

BMF

100

WHT 540

BMF

100

- - -

5 B I HOLDOWN

ANGLE

BRACKET

WHT

620

TTF 200

WHT

620

TTF 200

WHT

540

BMF 100

WHT

540

BMF 100

WHT

540

BMF 100

- - -

5 C R HOLDOWN

ANGLE

BRACKET

WHT

620

TTF 200

- - - - - - -

8 A R HOLDOWN

ANGLE BRACKET

WHT

620 TTF

200

WHT

620 TTF

200

WHT

620 TTF

200

WHT

620 TTF

200

WHT

620 TTF

200

WHT

540 BMF

100

WHT

540 BMF

100

WHT

540 BMF

100

8 A I HOLDOWN ANGLE

BRACKET

WHT 620

TTF

200

WHT 620

TTF

200

WHT 620

TTF

200

WHT 620

TTF

200

WHT 620

TTF

200

WHT 540

BMF

100

WHT 540

BMF

100

WHT 540

BMF

100

8 B R HOLDOWN ANGLE

BRACKET

WHT 620

TTF

200

WHT 620

TTF

200

WHT 620

TTF

200

WHT 620

TTF

200

WHT 620

TTF

200

WHT 540

BMF

100

WHT 540

BMF

100

WHT 540

BMF

100

8 B I HOLDOWN

ANGLE

BRACKET

WHT

620

TTF 200

WHT

620

TTF 200

WHT

620

TTF 200

WHT

620

TTF 200

WHT

620

TTF 200

WHT

540

BMF 100

WHT

540

BMF 100

WHT

540

BMF 100

8 C R HOLDOWN

ANGLE

BRACKET

WHT

620

TTF 200

- - - WHT

620

TTF 200

- - -

The number and the distribution of the

connection in the SFRS is calculated for

each examined configuration following the

procedures reported in Figure 2.

4.3. Design method

The earthquake action for these case study

buildings was calculated according to

Eurocode 8 (CEN, 2013) and the associated

Italian regulations (MIT 2008) using design

response spectra for building foundations

resting on ground type C*, assuming the

PGA equal to 0.35g (the highest value for

Italy) with a building factor of = 0.85.

[*Deep deposits of dense or medium dense

sand, gravel or stiff clay with thickness from

several tens to many hundreds of meters].

The seismic action was calculated starting

from the elastic spectra and applying an

initial q-reduction factor of 2 (CEN, 2013).

The coefficient kr was taken equal to 1.0 for

617

regular configurations and 0.8 for non-

regular configurations.

Figure 6 shows adopted design spectra, and

T1 values determined by simplified formula

and numerical frequency analyses methods

for configurations A R 3-5- 8.

Connections were first designed using the

force pattern obtained applying linear elastic

static analysis (CEN, 2013) and the seismic

action defined by taking T1 = T1_EC8.

Connection designs were then refined using

the rotation and translation force equilibrium

approach described by Gavric et al. (2011)

and Pozza and Scotta (2014) and the iterative

design process in Figure 2.

Figure 6. Design spectra and calculated periods according to CEN 2013

4.4. Finite element (FE) models

Numerical models of the investigated

building were realized using the finite-

element code Strand 7 (2005). The

illustrative FE model in Figure 7 uses linear

elastic shell elements to represent CLT

panels and link elements to simulate the

elastic stiffnesses of connectors. Beam

elements with pinned end conditions were

used to represent beam members

interconnecting perimeter shear walls and

shear walls in the building core at the top of

each storey.

Horizontal slabs elements in floor and roof

diaphragms were assumed to be rigid in-

plane. All the 15 building configurations

have been modelled respecting the

geometrical features and connection

stiffness’s in Table 2.

It is important to underline that the adopted

FE model is a limiting condition

representing the maximum deformability of

the system since the interaction between the

orthogonal walls and the out of plane

stiffness, provided by the interposed floor

slabs, are neglected.

On the other hand, FE models did not take

into account nonlinear deformability or large

displacements effects.

4.5. Finite element (FE) models

Following the iterative procedure reported in

Figure 2, the variation of calculated building

principal elastic periods (T1) among the

iterations was investigated by means of

modal response spectrum analyses of the

case study buildings. The procedure

converges in tree iterations for all examined

building configurations. Figure 8 reports

obtained values for each iteration of the

procedure. The alternative values given

represent effects of taking connection

stiffnesses (kconn) equal to values derived

from Eurocode 5 (kser) versus values derived

from experiments (ktest-method “a” / “b”). The

reference T1 values obtained using the

Eurocode 8 (CEN 2013) approach are also

reported.

0.0

0.1

0.2

0.3

0.4

0.5

0.0 0.2 0.4 0.6 0.8 1.0 1.2 1.4 1.6 1.8 2.0

Sd [

ag/g

]

T [s]

Design Spectra

3 A R - EC8

5 A R - EC8

8 A R - EC8

3 A R - kconn = kser



3 A R - kconn = ktest-meth. b




Figure 9 shows that the difference between

the predicted values of principal elastic

period (T1) at first iteration and at

convergence (delta) can be relevant and

spanning form 10% to 45% for the

investigated case study buildings. In addition

it is found that such error depends

substantially from two variables: (1) in

height building regularity and (2) number of

storey.

Regarding to the in height regularity (Figure

9) it results that the error “delta” for the non-

regular configuration (ie. A-B I 3-5-8) is

about 10% greater than he correspondent

regular configuration (ie. A-B-C R 3-5-8). It

means that the 1st temptative connection

distribution provided by approximated

Linear Static Analysis is acceptable only for

regular configuration and cannot be used for

non-regular configuration without relevant

error.

a) b)

c)

d) e) f)

Figure 7. View of the adopted FE model: a) 3AR configuration, b) 5AI configuration, c) 8BR

configuration, d) rigid story diaphragm detail, e) base connection detail and f) inter story

connection detail

Similarly the storey number affect the

reliability of the results obtained with the

preliminary Linear Static Analysis:

increasing the height of the building the error

“delta”, on the principal elastic period values

(T1), increases of about 10% switching from

3 storey to 5 storey and of 25% switching

from 3 storey to 8 storey.

Results in Figure 9 shows also that the

different values of stiffness assigned to

connections does not affect significantly the

variability of results among the iterations

still convergence.

4.6. Analysis results

Results presented here refer to the

convergence condition of the iterative

procedure and were obtained by modal

response spectrum analyses of case study

buildings, Figure 10 to 13. Those figures

show calculated building principal elastic

periods (T1), base shear forces (v) on angle

brackets at the Ultimate Limit State (ULS),

uplift forces on base hold-down anchors at

ULS (N), and the maximum inter-storey drift

619

values () at Damage Limitation State (DLS).

Figure 8. Calculated principal elastic periods (T1) for the three iterations of the procedure

Figure 9. Calculated free edge base uplift forces (N) for the three iterations of the procedure

The given alternative values represent effects

of taking connection stiffnesses (kconn) equal

to values derived from Eurocode 5 (kser)

versus values derived from experiments

(ktest). Figure 10 include also T1 calculated

with the simplified formula given by (CEN,

2013). Inter-storey drift was calculated for

each case study building using the Modal

Response Spectrum Analyses and the

Damage Limit State design spectrum.

Observing Figure 10 it is apparent that: (1)

in most cases use of experimental connection

stiffnesses (kconn = ktest) leads to much larger

T1 values than those predicted based

Eurocode 5 based estimates of connection

stiffnesses (kconn = kser); (2) using the simple

formula given by Eurocode 8 leads to low

estimates of T1 values. Interestingly use of

Eurocode 5 based estimates of kconn results is

estimates of T1 relatively close to simple

formula values; (3) “method a” of EN 12512

(CEN, 2006) provide stiffness values closer

to those form analytical prediction based on

Eurocode 5 (CEN, 2014) formula. However

3AR 3AI 3BR 3BI 3CR 5AR 5AI 5BR 5BI 5CR 8AR 8AI 8BR 8BI 8CR

1st - kconn. = kser 0.20 0.16 0.20 0.16 0.16 0.29 0.24 0.30 0.25 0.24 0.47 0.40 0.47 0.39 0.32

2nd - kconn. = kser 0.22 0.18 0.22 0.19 0.18 0.34 0.30 0.35 0.31 0.28 0.61 0.55 0.61 0.55 0.42

conv.- kconn. = kser 0.22 0.19 0.22 0.20 0.18 0.37 0.34 0.38 0.35 0.30 0.73 0.69 0.73 0.68 0.50

1st - kconn. = ktest_"meth. a" 0.29 0.22 0.28 0.22 0.21 0.42 0.30 0.40 0.30 0.28 0.63 0.49 0.61 0.46 0.35

2nd - kconn. = ktest_"meth. a" 0.31 0.25 0.30 0.25 0.23 0.49 0.38 0.46 0.37 0.32 0.82 0.68 0.78 0.64 0.45

conv. - kconn. = ktest_"meth. a" 0.32 0.27 0.30 0.26 0.23 0.54 0.44 0.51 0.42 0.35 0.98 0.84 0.93 0.81 0.53

1st - kconn. = ktest_"meth. b" 0.42 0.32 0.40 0.30 0.30 0.63 0.41 0.57 0.40 0.34 0.91 0.64 0.82 0.60 0.42

2nd - kconn. = ktest_"meth. b" 0.45 0.36 0.43 0.34 0.32 0.72 0.52 0.66 0.50 0.41 1.16 0.90 1.07 0.85 0.55

conv. - kconn. = ktest_"meth. b" 0.47 0.39 0.44 0.37 0.33 0.80 0.61 0.73 0.57 0.45 1.40 1.14 1.30 1.08 0.66

Eurocode 8 0.26 0.26 0.26 0.26 0.26 0.38 0.38 0.38 0.38 0.38 0.54 0.54 0.54 0.54 0.54

0.0

0.2

0.4

0.6

0.8

1.0

1.2

1.4

1.6

Pri

nci

pal

vib

rati

on

per

iod

[s]


delta - T1_kconn. = kser 9.0% 18.0% 7.0% 18.0% 11.0% 21.0% 30.3% 21.8% 29.5% 20.1% 35.2% 42.6% 36.3% 42.6% 35.2%

delta - T1_kconn. = ktest_"meth. a" 8.0% 19.0% 8.5% 17.0% 9.0% 22.7% 31.2% 22.2% 28.6% 19.3% 35.6% 41.6% 35.2% 43.3% 36.0%

delta - T1_kconn. = ktest_"meth. b" 11.0% 18.0% 10.0% 20.0% 10.0% 21.7% 32.0% 23.0% 29.5% 23.5% 35.2% 44.0% 37.0% 44.0% 37.0%

0%

5%

10%

15%

20%

25%

30%

35%

40%

45%

50%

va

ria

bil

ity o

f p

rin

cip

al

ela

stic

per

iod


results suggest that neither of those

approaches are reliable ways of estimating

principal natural periods of buildings having

SFRS consisting of CLT cores and perimeter

shear walls.

Figure 10. Predicted principal elastic periods (T1)

Consequences of discrepancies in kconn

values from those found by testing varied in

their effects on v, N and values, but, in

general, results show that the method used to

estimate the connection stiffnesses can alter

design force and lateral drift estimates by

substantial amounts, Figure 11 to 13.

Figure 11. Predicted base shear per unit of length (v)

Figure 12. Predicted free edge base uplift forces (N)


T1_Eurocode 8 0.26 0.26 0.26 0.26 0.26 0.38 0.38 0.38 0.38 0.38 0.54 0.54 0.54 0.54 0.54

T1_kconn. = kser 0.22 0.19 0.22 0.20 0.18 0.37 0.34 0.38 0.35 0.30 0.73 0.69 0.73 0.68 0.50

T1_kconn. = ktest_"method a" 0.32 0.27 0.30 0.26 0.23 0.54 0.44 0.51 0.42 0.35 0.98 0.84 0.93 0.81 0.53

T1_kconn. = ktest_"method b" 0.47 0.39 0.44 0.37 0.33 0.80 0.61 0.73 0.57 0.45 1.40 1.14 1.30 1.08 0.66

0.0

0.2

0.4

0.6

0.8

1.0

1.2

1.4

1.6

1st

vib

rati

on

pe

rio

d [

s]


N_kconn. = kser 128.1 170.9 152.0 178.1 161.3 316.0 420.6 355.4 447.3 373.2 511.6 884.1 559.5 759.7 897.0

N_kconn. = ktest_"method a" 122.6 164.0 138.6 188.7 137.9 258.6 379.2 283.1 415.4 340.4 389.2 649.2 413.5 586.9 724.9

N_kconn. = ktest_"method b" 138.5 185.7 149.4 232.1 138.5 246.2 403.8 260.1 455.7 366.9 334.4 527.3 339.5 516.1 678.8

0

100

200

300

400

500

600

700

800

900

1000

bas

e u

plif

t fo

rce

[K

N]


v_kconn. = kser 28.20 33.20 32.70 40.30 35.80 42.30 51.70 51.80 64.20 50.40 50.20 64.60 61.60 82.60 79.70

v_kconn. = ktest_"method a" 25.90 31.33 30.41 35.28 29.90 33.12 46.46 41.49 56.40 46.41 38.73 51.52 47.98 73.09 65.00

v_kconn. = ktest_"method b" 28.10 34.90 33.40 36.40 29.20 29.70 49.30 38.40 58.40 50.50 34.00 47.40 42.70 76.30 61.60

0

10

20

30

40

50

60

70

80

90

bas

e s

he

ar [

KN

/m]

621

Figure 13. Predicted maximum inter-storey drift ()

Comparing results of Figure 11 (base shear

force) with those of Figure 12 (base uplift

force) it is possible to observe that the

variability of base shear force values induced

by different stiffnesses estimation is lower

than that of base uplift force values

especially for the 5 and 8-storey

configuration. This means that for high-rise

buildings the lateral deformability is mainly

controlled by holdown stiffness that define

the rocking behaviour of shear walls.

Otherwise the sliding effects is reduced and

the seismic horizontal action is uniformly

distributed among the angle brackets

connections.

The predicted free edge base uplift forces

provided by the estimation of connection

stiffness according to code (i.e. Kconn = kser)

are much greater than those given by test

stiffness estimation (i.e. Kconn = ktest)

particularly for 8-storey configuration. A

direct consequence of this is that using

analytical estimation of connection stiffness

induces an overdesign of holdown

connections for high-rise CLT buildings. On

the contrary, for low- and mid-rise building

configuration the effect of connection

stiffness estimation on base shear and uplift

forces is small and stable among the

different examined configurations.

Another aspect that can be deduced form

obtained results shown in Figure 11 and 12

is that the non-regular configuration are

characterized by a greater susceptibility to

the variation of connection stiffness

estimations.

Estimates of predicted values of inter-storey

drift () reported in Figure 13 are really

sensitive to connection stiffness estimation

particularly for eight-storey buildings.

As results inter-storey-drift was estimated to

be up to four times larger assuming kconn =

ktest-meth. “b” than assuming kconn = kEC5. Inter-

storey drift values provided by connection

estimation based on experimental method

“a” (i.e. kconn = ktest-meth. “a”) appear to be the

most suitable approach for a reliable

estimation of lateral deformation of the

SFRS.

5. Conclusions

Results obtained in this work demonstrate

that hold-down and shear connections used

to assemble and connect CLT wall panels

largely determine the behaviors of SFRS. It

is therefore crucial to properly represent the

stiffnesses of connections during structural

analyses from which principal elastic period,

peak dynamic forces flowing through wall

and connection elements and inter-storey

drift are estimated.

A specific design procedure suitable for an

efficient and safe design of mid- and high-

rise CLT buildings is proposed basing on

reliable definition of the connection stiffness

via code provisions and experimental tests.

The main feature of this procedure consists

in the iterative approach necessary to define

the best connection arrangement among the

shear walls and therefore the reliable

estimation of the building global stiffness.


d_kconn. = kser 2.9 1.80 2.50 1.90 1.40 5.60 4.40 5.10 4.30 2.60 9.90 8.80 8.80 7.80 4.70

d_kconn. = ktest_"method a" 6.99 4.42 6.16 4.00 2.39 9.75 6.49 8.79 6.03 3.54 14.35 10.44 12.56 9.29 4.74

d_kconn. = ktest_"method b" 12.30 7.80 10.90 6.80 3.80 15.60 9.70 14.00 8.80 5.10 21.30 13.90 18.50 12.40 5.60

0

5

10

15

20

25

max

inte

r st

ore

y d

rift

[m

m]


The iterative process start with a preliminary

design of connection systems based on code

provisions both for principal elastic period

value and connection’s parameters. Then the

connection distribution can be iteratively

refined referring to more precise analyses for

the estimation of the principal elastic periods

and therefore of the action on the connection

elements. The procedure also include a

verification of the interaction between the

tensile and shear force or displacement

according to specific resistant domain

prescribed by standards.

The crucial aspect of the proposed procedure

consists in the choice of the best approach

for a reliable estimation of the connection

elastic stiffness. In this work two different

approach are used: the first one is analytical

and based on Eurocode 5 (CEN, 2014)

provisions while the second one refer to the

experimental tests and to the stiffness

estimation provided by EN 12512 method

“a” and “b” (CEN, 2006).

Validation of proposed procedures on 15

different case study building configurations,

characterized by different number of storeys

and connection arrangement, demonstrate

that the difference between the predicted

values of principal elastic period at first

iteration and at convergence is relevant

(form 10% to 45%) and depends

substantially from in height building

regularity and number of storey. Procedure

provided by codes sounds suitable only for

building up to 3-storey characterized by in

plant and in height regular shear walls and

connections distribution.

Regarding to the effect of stiffnesses

estimation, case studies suggest that

principal elastic periods values

underestimated by up to 50 percent is a

realistic scenario unless designers use test

data to estimate connection stiffnesses.

Large errors occurring during subsequent

calculation of shear and hold-down forces

and inter-storey drift is also highly feasible.

Finally results shows that for buildings

having three to eight storeys principal elastic

period estimates, shear and uplift forces at

bases of wall panels, and inter-storey drift

can all be miscalculated by substantial

margins if a standardized procedure for

calculating and schematizing of connections

behaviour is not followed.

Acknowledgments: This research, as part of

the AdriaHub project, was supported by the

European Union (EU) inside the Adriatic

IPA CBC Programme, an Instrument for the

Pre-Accession Assistance (IPA) of neighbour

countries of Western Balkan thanks to

investments in Cross-Border Cooperation

(CBC). Details are available in (Savoia et al.,

2016).

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Polastri A. Trees and Timber Institute,

National Research Council

of Italy (CNR IVALSA)

Via Biasi 75, 38010 San

Michele all’Adige (TN)

Italy

[email protected]

Pozza L. University of Bologna,

Dept. of Civil, Chemical,

Environmental and Materials

Engineering

Italy

[email protected]



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