ConnectionsSpecial attention should be paid to the design of connections in any structural design because connections are responsible for the majority of structural failures. Furthermore, the cost associated with constructing connections can take up a large chunk of a project’s funds. Good connection design “is all about following load through all the elements in its path” (Fadden, 2008). For our hotel design this meant considering multiple load paths, since we have a large variety of structural connections. Besides insuring that the connections are strong enough to transfer loads from one member to another, they must also be economical and constructible. Methods to increase economy and constructability in this project included: (1) avoiding field welding when possible, by having the connecting parts shop welded or field bolted, (2) uniformity in connection design, and (3) avoiding complete-joint-penetration (CJP) welds whenever possible, in favor of partial-joint-penetration welds (PJP) or fillet welds. All structural connections can be classified into one of three categories: simple, fully-restrained, or partially restrained. Simple connections allow uninhibited rotation of the connecting elements and are usually classified by the largest force that they resist. In the case of gravity only beam-to-column connections they are classified as “shear connections”. Fully-restrained connections allow negligible rotation of the connecting elements and transfer moment; while easy to model in structural analysis software, in practice fully-restrained connections are difficult to design due to the negligible rotation requirement. Partially-restrained connections allow rotation that can’t be ignored in the analysis, and transfer moment. Correct classification of each connection design is important in order to accurately model the connection in the analysis.

Simple Shear ConnectionsMembers are not a part of the SLRS were connected using simple shear connections. The use of simple shear connections was adequate because these frames were only called upon to resist gravity loads. The simple shear connection chosen for this project is a single clip angle which is to be shop welded to the column flange (or web, depending on orientation) and field bolted to the beam web. The advantages of this type of simple shear connection is that it is easily field installed and can accommodate multiple orientations, such as a beam framing into the column web, beam framing in the column flange, or a joist framing into a girder. The limit states checked for this connection included: (1) bolt shear, (2) bolt bearing, (3) shear yielding, (4) shear rupture, and (5) block shear.

The worst case scenario for a simple connection was on the first floor, where a W14X74 beam was framing into the flange of a W14X342 column. Based on a linear analysis of the structure with all gravity loads applied, the shear at the connection was 86 kips. Plugging this data into our single angle shear connection spreadsheet an adequate single angle connection was found. The connection calls for a L5X5X7/8 angle with 3, A490X bolts attaching one leg to the beam web and L-shaped fillet welding with a thickness of ½ in and a length of 10.5in connecting the other leg to the column flange. A detail and rendering of the connection can be seen in Figure 1 and Figure 2 respectively. The input information and analysis of the connection can be seen in Appendix A. A similar connection with the same members and forces was also designed to frame into the column web. This connection can be seen in Figure 3. The accompanying input data and analysis can be seen in Appendix B.

Figure 1 Beam-Column Flange Detail

Figure 2 Shear Connection Rendering (Beam-Column Flange)

Figure 3 Shear Connection Rendering (Beam-Column Web)

Moment Frame ConnectionThe SMF connections for our structure were designed in accordance with AISC 341 and AISC 358. The moment resisting connection chosen was the bolted extended end-plate moment connection. The bolted extended end-plate moment connection comes in three configurations the four-bolt unstiffened (4E), the four-bolt stiffened (4ES), and the eight-bolt stiffened (8ES). The design requirements for this

connection can be found in AISC 358 Chapter 6. Typical configurations of this connection can be seen in Figure 4.

Figure 4 Extended End-Plate Moment Connection 4E, 4ES, and 8ES (AISC 358: American Institute of Steel Construction, 2010)

The extended end-plate moment connection was chosen above various other prequalified moment connection in AISC 358 because it has the widest range of applicability to our structure, which allowed for more uniformity. Also, this connection has an ample design guide (AISC Design Guide 4 Extended End-Plate Moment Connections) which allowed an accurate spreadsheet to be designed for all three configurations. Furthermore, this connection is classified as fully-restrained, which makes analysis of the structure easier. The limitations on which extended end-plate configuration is applicable to various connections can be seen in Figure 5. One of the three configurations was applicable to all of the moment-resisting connections in our structure. Selection of a suitable connection for each SMF connection can be seen in Figure 6. Great effort was taken to avoid the use of column doubler plates due to their high cost of installation. When a certain connection necessitated a doubler plate, the cost of installing a doubler plate was compared to the cost of increasing the weight of the column. In most cases beefing up the weight of the column was the more economical choice. Also, beefing up the columns gave the moment frames better strong-column-weak-beam behavior. The protected zone for stiffened extended end-plate moment connections “is the portion of the beam between the face of the column and a distance equal to the location of the end of the stiffener plus one-half the depth of the beam or 3 times the width of the beam flange, whichever is less” (AISC 358: American Institute of Steel Construction, 2010). In accordance with AISC 341, no shear studs were applied within the protected zones of the connections.

Figure 6 Selection of Extended End-Plate Configurations

The worst case scenario for a moment resisting connection in our structure involved framing a W21X83 beam into a W14X550 column. The connection contained 1,400 kips of axial load, 241 kips of shear in the beam, and 428 kips of shear in the column immediately above the connection. The required flexural resistance of the connection was derived by applying AISC 341 Eq. (9-1). Based on the sizes of the members and the necessary strength an 8ES moment connection was designed to connect the two members. Using a spreadsheet the connection was checked for many limit states including: (1) flexural yielding of the end plate, (2) yielding of the column panel zone, (3) tension rupture of the end plate

bolts, (4) shear rupture of the end plate bolts, and (5) rupture of various welded joints. A detail of the connection can be seen in Figure 7. Also, the input data and analysis for

the connection can be seen in Appendix C. While continuity plates were not necessary for this connection based on strength requirements, they were added based on best engineering practices. The final connection which is in conformance with all of the pre-qualifications outlined by AISC 358 is able to undergo an inter-story drift angle of .06 radians, which is above the requirement outlined in AISC 341 9.2a of .04 radians.

Figure 5 Limitations of Extended End-Plate Configurations

Figure 7 Detailed 8ES Extended End-Plate Moment Connection

Special Concentrically Braced Frame ConnectionsSCBF connections must be designed to resist the full compression, tension, and flexural forces of their connecting members even once those members become inelastic. The maximum tension force equates to the yielding of the braced members which can be derived from Equation 1, where Ry is a material over-strength factor. The required compressive strength is governed by the buckling limit state of the brace. The required flexural strength of the connection is derived with Equation 2. Since the brace will experience buckling before tension yielding, an imbalance of forces is created at the bracing connections. Gusset plates are used to transfer these forces to the beam and column. The distribution of forces through the gusset plate was calculated using the Uniform Force Method. The spreadsheet that was used to design our SCBF connections, which can be seen in Appendix D was downloaded from SteelTools via AISC.org. An example SCBF connection detail can be seen in Figure 8. This detail was designed for the SCBF that contained the largest braces, which meant the connections had to resist the largest required forces. As can be seen from Figure 8, the connections required the gusset plate to be PJP welded to the beam and column using a ¾” throat. The brace member which was a square HSS was slotted and then PJP welded to the gusset using a 3/16” throat. The gusset plate required a thickness of 1.5” in order to resist the governing limit state of tension yielding. Other limit states checked for the gusset plate included: block shear, buckling, shear rupture and weld strength. Limit states checked for the beam and column included: web yielding and web crippling. The beam-to-column connections in SCBF’s can be designed as simple connections, so we designed them using the same simple shear connection as we did on the gravity-only frames.

F=RyFyAg

Equation 1 Tension Yielding Force of Brace (AISC 341:American Institute of Steel Construction, 2010)

F-=1.1RyMp

Equation 2 Equation for Required Flexural Strength of SCBF Connection

Figure 8 SCBF Connection Detail

Eccentrically Braced Frame ConnectionsEBF’s ideally isolate all of their inelastic deformation in the shear link. Thus, there are many requirements outlined by AISC 341, which are meant to assure that the EBF connections are strong enough to endure the large forces induced by this inelastic behavior. The required strength of bracing connections in an EBF are not as large as those for SCBF, because unlike in a SCBF, the braces in a EBF are designed to remain elastic. The bracing connections for the EBF’s were designed using the same spreadsheet as the SCBF, but with the required strength reduced. While technically not a connection because the beam in an EBF is continuous, the shear link in an EBF requires special detailing in order to withstand the necessary inelastic deformations. One requirement for the link is that “full-depth web stiffeners must be provided on both sides of the link web at the diagonal brace ends of the link” (AISC 341:American Institute of Steel Construction, 2010). Also, intermediate web stiffeners are required depending on the cross-sectional properties and length of the link. The link designed for this report, which can be seen in Figure 9 required 5 intermediate stiffeners. The spreadsheet used to design the shear link can be seen in Appendix E. Connection of the beam to the column in EBF’s where the link is in the middle of the beam can either be designed as pinned or moment-resisting. We chose to design this

connection as pinned, which allowed it to me modeled the same as the gravity-only beam-to-column connections.

Figure 9 EBF Link Detail

Column Base PlatesBase plates help transfer the loads from the superstructure to the foundation. Base plates must allow the system that they support to exhibit the ductile behavior that it was designed for. In order to do this AISC 341 outlines strength requirements that the base plate must meet. The requirements include: (1) axial strength, which is computed from the column required strength in combination with the vertical component of the connection of any braces present, (2) shear strength, which is computed from a mechanism in which the column forms plastic hinges at the top and bottom of the first story, in combination with the horizontal component of the connection required strength of any braces present, and (3) flexural strength , which is computed from a mechanism in which the column forms plastic hinges at the base plate, in combination with the required flexural strength of any braces present. Like other connections, base plates can be designed as simple (pinned), partially-restrained, or fully-restrained. For our system, we chose to design the base plates as pinned because the cost and difficulty of providing a full-restrained base plate connection would be extremely high. The most extreme base plate connection involved a W14X655 column that had 3,600 kips of axial load and 500 kips of shear. The detail of this base plate connection can be seen in Figure 10. The full analysis and design of this

connection can be seen in Appendix F. The column is attached to the plate via CJP welds.

Figure 10 Base Plate Connection Detail

Shear Wall Connection DesignUsually the beams connect to the core shear wall through a shear connection. Although this connection provides some moment resistance, it is generally accepted it is of a negligible magnitude. The main design issues include: (1) the connection between the steel beam and shear tab which is welded onto the embedded plate and (2) the transfer of forces of gravity loads and diaphragm forces to the wall. Special attention was given to the fact that the diaphragm forces could be either tensile or compressive. The general design method for shear wall connections as given by the Handbook of Structural Steel Connection Design and Details by Akbar R. consist of:

1. Based on an assumed layout of studs, establish the tensile capacity as the lesser of strength of the stud or concrete core.

2. Assuming that all the applied shear is resisted by the studs in the compression region; calculate the required number of studs. The shear capacity is taken as the smaller of shear capacity of a single stud.

3. Using the stud arrangement obtained in step 2, compute tensile capacity of the stud group.4. Increase the value of Tu by 50%to ensure adequate ductility.5. Calculate the depth of compression region, kd:

Equation 3

6. Calculate the required depth of the embedded plate

Equation 4

Equation 5

7. Check the capacity of studs under combined actions of tension and shear.

From the ETABS model we have the critical shear and axial load of the beams connecting to the shear wall core, which is Tu=29.25kips and Vu=22.4kips. According to the design procedure above, we calculated the tab section to have a length of 6in and a width of 4in., the plate to have a depth of 8in, a width of10in and a thickness 0.5in, and 0.75in anchors embedded in the concrete. The detail of the connection can be seen in Figure 11 and Figure 12.

Figure 11 Detail of shear connection between steel beam and RC wall 1

Figure 12 Detail of shear connection between steel beam and RC wall 2

Column SplicesColumn splices are necessary in structures due to the limitations on transportation and erection of columns. If it were possible to transport and erect 25 story columns than alas column splices would be unnecessary on this project. Until that day, attention should be paid to the ample design of column splices. The purpose of column splices is to fully transfer whatever forces are acting in one column to another. In order to achieve this goal, AISC 341 outlines a few requirements for column splices. One requirement that applies to all columns in a building that employs some type of SLRS is that column splices must be placed 4ft away from the beam-to-column connection. This requirement is intended to limit the forces that the splice is required to resist by moving the splice away from the plastic region of the beam-to-column connection. Another requirement that applies to all splices in a building are that they must be capable of resisting a shear force equal to the plastic flexural strength of the smaller column divided by the story height. There are further strength requirements for column splices depending on what type of SLRS they are a part of. In our structure, column splices are provided every five floors. Since the typical story height is 12ft this means that a typical column in our structure will need to be shipped in lengths of 64ft (= (12x5) +4). In our structure, we have columns that function as members in gravity-only systems, special concentrically braced systems, moment-resisting systems, eccentrically braced systems, or a combination of these. Each of these systems comes with different design requirements. If a column is a member in multiple SLRS’s than it must be designed under the most stringent requirements. The splices used in our design are all-bolted cover plate splices. These were chosen because they are acceptable under all requirements and they can account for differences in the geometry of the joining columns. Bolted connections were used instead of welding due to the added difficulty of constructing CJP welds to thick members. A typical configuration of this column spice can be seen in Figure 13.

Figure 13 All-Bolted Cover Plate Column Splice (Joints in Steel Construction: Simple Connections, 2002)

The most extreme column splice occurrence in our structure involved a W14X655 and W14x550 columns that were a part of a special concentrically braced system. The forces acting through the splice were extracted from SAP2000, but turned out to be smaller than the required design forces, as specified by AISC 341. So, the splice was designed with the AISC 341 required forces. The final splice design required 48, A490 bolts. The bolts were designated as slip-critical, as specified by AISC 341 for splices in SLRS’s. Web and flange packs were required for the splice in order to accommodate the different web and flange thicknesses of the columns. A division plate was not required because the depth of the columns were similar enough to one another. Since the splice is bearing, the bolt design was controlled by the required shear resistance in the splice. The detailed splice can be seen in Figure 14.

Figure 14 Detailed Moment Splice

Moment-Resisting FramesIn our preliminary report, one of the four designs included a SLRS that utilized only moment resisting frames. After analysis of this design under the prescribed seismic and wind loads it was evident that this system was too ductile for the height of the building and the magnitude of the lateral loads. To combat this issue, our final design utilizes stiffer elements such as SCBF’s and shear walls, with SMF’s acting as a back-up system. Since our final SLRS incorporates a dual system, the moment frames in accordance with ASCE 7-10 were designed to be capable to withstand 25% of the design seismic forces. Similar to the process used in the preliminary design, we began the design of the moment frames by using approximate methods such as the portal method. The initial member sizes were chosen based on the forces derived from the approximate methods. The next step in the process was to model the SMF’s in ETABS. The SMF’s needed to be modeled and fine- tuned in a separate model from the rest of structure in order to make sure they were capable of resisting 25% of the lateral load. Using the steel design feature, the moment frames were designed in accordance with AISC 341. Once the SMF’s were designed to resist 25% of the lateral load they were placed in the full structure. In order to check the validity of the ETABS design function, the column-beam moment ratio of single SMF configuration was calculated. An example of the hand calculation can be seen in Equation 6. The configuration involved 2 W21X111 beams framing into a W14X655 column. There was one other beam also framing into the column, but it was only a part of the gravity-system, so it was not included in the equation. The equation resulted in a

ratio of 3.17, which is way above the requirement of 1.0. Based on this high ratio, it can be concluded that the plastic hinges will in fact occur in the beams.

Equation 6 Column-Beam Moment Ratio (AISC 341:American Institute of Steel Construction, 2010)

Shear Wall Design

IntroductionIn our design project, the shear walls play a major role in the seismic load resisting system (SLRS). Unlike the braced systems which are evenly dispersed in great numbers throughout the building, the shear walls are only located in the center of both legs of the building. The reasons for placing the shear walls in the middle of legs include: (1) to minimize torsional effects by keeping the center of rigidity and the center of mass closer together, and (2) to maintain architectural freedom in the building, which is accomplished by only having shear walls wrapped around the elevators. In the long building, we have two E shaped shear walls forming a RC shear wall core. The short building contains two C shaped shear walls. The location of these shear walls can be seen in Figure 15. The reinforced concrete shear wall cores are connected to each other by steel beams. The steels beams utilize simple connections, which only transfer shear. The advantage of connecting the shear wall cores is that they become stiffer, thus increasing their lateral load resistance. The disadvantage of integrating the shear walls is they must be thicker and longer than if they were individual shear walls.

Figure 15 Location of Shear Walls

Stiffness of shear wallThe first step in the design of the shear walls is to find the relative stiffness of the shear walls. The relative stiffness determines how much of the lateral load the shear walls will attract.

Usually shear walls are categorized as either squat shear walls or slender shear walls. Short, one- or two-story shear walls generally can be designed using a strut-and-tie model (Figure 17). If the wall is more than 3 or 4 stories in height, lateral load is resisted mainly by flexural action of the vertical cantilever wall, rather than by in-plane strut-and-tie forces (Figure 17). Obviously in our project which consists of 25 stories, the shear walls were designed as a vertical cantilever wall. According to the stiffness of cantilever beam equation which considers both flexure and shear, the lateral displacement can be calculated using Equation 7.

Equation 7 Cantilever Beam Equation

Where

q = the function of lateral forces

H= the height of the shear wall

E = Young’s modulus

I= Moment of inertia

μ= cross section factor

G = shear modules, for concrete G=0.4E

A = cross section areaΔ

Figure 16 Slender shear wall stiffness calculation model

In our case we simplified the analysis by applying the load to the top of the shear wall, so the lateral displacement at the top can be calculated using Equation 8.

Equation 8 Shear Wall Displacement at the Top

Where,

F = the lateral force at the top of the shear wall

Figure 17 Squat shear wall strut-and-tie, and slender shear wall design model

Based on the stiffness we obtained from Equation 8, we assume the shear walls are going to take about 25%-30% of the lateral loads. For the long building this consisted of 34,082kips-ft in the W-E direction and 22,888kips-ft in the N-S direction. For the short building the load was 20,480kips-ft in the W-E direction and 91,96kips-ft in the N-S direction. Each E and C shape shear wall takes half of the lateral loads above in their own part.

Design for flexural strengthBased on the height and the lateral forces, we assumed the cross sections of the shear walls to be as follows, each of the E shape shear walls in the long building contain a web that is 11.5ft long and has a width of 1.5ft, the flange is 17ft long and also has width of 1.5ft. The two are symmetric with respect to the X axis. The web of the C shaped shear walls in the short building are 11.5ft long with a width of 1.5ft, the flange is 7ft long and has a width of 1ft, Both C-shaped shear wall are symmetric about their y-axis. The details for the E-shaped and C-shaped shear walls are presented in Figure 18 and Figure 19, respectively.

Figure 18 Long building E shape shear wall in N-S view

Figure 19 Short building C shape shear wall in W-E view

Once the dimensions, design moment and axial loads applied to the centroidal axis of the cross section are known the vertical reinforcement can be determined. The design approach for the shear walls followed the procedures outlined in Seismic Design of Reinforced Concrete and Masonry Buildings by Paulay T. Assuming the constituent wall segments, such as 1, 2, and 3 in Figure 20, the amount of reinforcement in segment 2 is usually nominated and it often corresponds to the minimum recommended reinforcement as specified by ACI 318.

Figure 20 Wall section example (Seismic Design of Reinforced Concrete and Masonry Buildings of Paulay T.)

However, this assumption does not need to be made because any reinforcement in area 2 in excess of the minimum is equally effective and hence will correspondingly reduce the amounts required in the flange segments of the wall. By assuming that all bars in segment 2 and 3 will develop yield strength, the total tension force T2 and T3 can be found. Next, we may assume that when Ma=eaPa, the center of compression for both concrete and steel force C1 is in the center of segment 1. Hence the tension force required in segment 3 can be estimated from Equation 9.

Equation 9 (Seismic Design of Reinforced Concrete and Masonry Buildings of Paulay T.)

Equation 10

And thus the area of the reinforcement in this segment can be found. Practical arrangement of bars of bars can now be placed. Similarly, the tension force in segment area 1 is estimated when Mb=ebPb from

Equation 11

Equation 12

Further improvement with the estimates above may be made, if desired, by checking the intensity of compression forces. For example, when Pa is considered, we find that

Equation 13

And hence with the knowledge of the amount of reinforcement in segment 1, to provide the tension force T1, which may now function as compression reinforcement, the depth of concrete compression can be estimated

Equation 14

After the reinforcement has been chosen, the moment resistance of the wall should be computed by using the strain-compatibility analysis. The factored nominal moment capacity of the wall should be

Equation 15

Based on the approach above, firstly we split the E shape shear wall into two L and one T segments in N-S direction to consider the effect of the shear wall flange. The effective flange widths selection can be found from ACI Sections 8.10.2 and 8.10.3. In regions subject to earthquake, ACI Section 21.7.5.2 limits the flange width to the smaller of

(a) half the distance to an adjacent web and(b) 25 percent of the total height of the wall above the section under consideration.

In the L shape segment, the web is 10ft long, the length of effective flange is 5ft and so is the T shape segment. The dimension details of the segments of E and C shape shear walls are in Figure 21. But we will design the E shape shear wall in W-E direction and C shape shear wall in N-S direction as a whole shear wall due to the symmetry property.

Figure 21 Long building E shape shear segments in N-S direction

Following the flexural design approach above, we can get the detail reinforcement and flexural capacity of each segment. The reinforcement of critical section at the bottom of the shear wall is in Table 1. Notice that the reinforcement is compatible with two directions.

Table 1 Design of Flexural strength of shear walls

Long building N-S Direction L segment shear wall

Web Flange

bar No. 6 6

spacing(in) 18 18

Layers of Reinforcement 2 2

No. of bars 8 5

As(in^2) 7.04 4.4

Mu/Φ (k-ft) 3740

Mn(k-ft) 16075.12438

Long building N-S Direction T segment shear wall

Web Flange

bar No. 6 6

spacing(in) 18 18

Layers of Reinforcement 2 2

No. of bars 8 7

As(in^2) 7.04 6.01

Mu/Φ (k-ft) 5236.666667

Mn(k-ft) 27233.47984

Long building W-E Direction E shape shear wall

Web Flange1 Flange2 Flange3

bar No. 6 6 6 6

spacing(in) 18 18 18 18

Layers of Reinforcement 2 2 2 2

No. of bars 13 8 7 8

As(in^2) 11.44 7.04 6.16 7.04

Mu/Φ (k-ft) 18934.44444

Mn(k-ft) 82038.98744

Short building N-S Direction C shape shear wall

Web Flange

bar No. 6 6

spacing(in) 18 18

Layers of Reinforcement 2 2

No. of bars 8 16

As(in^2) 7.04 14.08

Mu/Φ (k-ft) 4598

Mn(k-ft) 17720.84706

Short building W-E Direction L segment shear wall

Web Flange

bar # 6 6

spacing(in) 18 18

Layers of Reinforcement 2 2

# bars 8 5

As(in^2) 7.04 4.4

Mu/Φ (k-ft) 5120

Mn(k-ft) 13948.6719

Design of shear strengthIn the slender shear wall design, the overturning moment controls the design; however we still need to place transverse reinforcement to integrate all the rebar forming the cage. This will improve the integrity and effectiveness of the shear wall. The nominal shear strength of the shear wall shall not exceed Equation 16.

Equation 16 Maximum Nominal Shear Strength

Where

Acv = shear area

Ρn = the ratio of shear reinforcement

This is equivalent to Vn=Vc+Vs, where Vc and Vs are the shear strength capacity of the concrete and reinforcement, respectively. We can get the transverse reinforcement in web detail in Table 2.

Table 2 Design of shear strength of shear walls

Long building N-S Direction E shear wall

L Segment T Segment

bar No. 9 8

spacing(in) 15 11

ρh 0.002194444 0.001994949

Vn (kips) 495.22 474.54

Base shear (Vu, kips) 152.24 212.88

Long building W-E Direction E shear wall

bar No. 8

spacing(in) 11

ρh 0.001994949

Vn (kips) 653.00

Base shear (Vu) 517.36

Short building N-S Direction C shear wall

bar No. 8

spacing(in) 11

ρh 0.001994949

Vn (kips) 949.07

Base shear (Vu) 226.06

Short building W-E Direction C shear wall

bar No. 8

spacing(in) 10

ρh 0.002633333

Vn (kips) 450.60

Base shear (Vu) 226.06

Figure 22 Long building reinforcement detail of flexural design by handwork

Figure 23 Short building reinforcement detail of flexural design by handwork

ETABS Design of shear wallAlong with hand calculations, we also modeled the shear walls into ETABS. ETABS can perform a finite element analysis, which is more precise than hand calculations because it takes into account more factors. However, modeling of the shear wall must be handled with caution because unless you really screw something up ETABS will still spit out an answer. It is up to the engineer to judge whether or not the answers are legitimate. For our design, we checked the ETABS results with our hand calculations to determine their accuracy.

From the ETABS design of the shear wall, we derived the required reinforcement which can be seen in Figure 24 and Figure 25. All shear walls use #7 bars with 12in spacing; the ratio of the reinforcement is 0.0067.

Figure 24 Long building reinforcement detail of flexural design by ETABS

Figure 25 Short building reinforcement detail of flexural design by ETABS

Appendix A

Appendix B

Appendix C

Appendix D

Appendix E

Appendix F

Appendix G

References