Moment Resisting Wood Post-To-Concrete Peir Connection

Moment Resisting Wood Post-to-Concrete Pier Connection

Biological Systems Engineering Department

University of Wisconsin-Madison

Date of Graduation: May 13, 2006

Paper Submission Date: May 22, 2006

__________________ ______ __________________ ______ __________________ ______

Aaron Flouro (Date) Kyle Bunnow (Date) Dr. David Bohnhoff (Date)

(Designer) (Designer) (Advisor)

Abstract:

The vast majority of agricultural structures are post-frame buildings. The posts in thesestructures are either connected to a concrete slab foundation, or they are embedded in theground in which case they are a main element of the building’s foundation. To minimizedecay, embedded wood posts must be treated with a wood preservative. Because the useof preservative-treated wood is somewhat problematic, it is advantageous to usesomething other than the standard embedded wood post. One such alternative is toreplace the embedded portion of a wood post (i.e., the only portion that typically needs tobe preservative-treated) with a concrete pier. The goal of this research project was todevelop the connection needed to adequately attach a concrete pier to a wood post. Morespecifically, the goal was to design a connection that would have a bending strength andstiffness near that of the components it was connecting. The approach to this projectinvolved breaking the connection into three main elements for design, theoretical analysisand laboratory testing: (1) a steel bracket portion, (2 ) a steel bracket-to-concrete pierconnection, and (3) a steel bracket-to-wood post connection.

For the attachment of a 3-layer, mechanically laminated post fabricated from nominal 2-by 6-inch lumber, the final proposed design consists of two 14-inch long by 5.5-inchwide plates that are 0.1875 inches thick. These plates are sandwiched between thelumber plies, and held in place with two _” bolts and four _” by 2 _” steel piercingscrews. The plates extend 2 inches into the concrete pier. Two rebars, each with a 180-degree bend, and a formed length of 24 inches are welded to the steel plates. The two180-degree rebars are joined by transverse rebar welded on to form a location for thelongitudinal rebar that will be used as reinforcement in the concrete pier.

Key Words:

Post-Frame, Wood-to-Concrete Connection, Post, Pier, Wood Preservative, MomentConnection, Mechanically Laminated Post, Mechanical Connection, Connection

Acknowledgements:

We would like to thank Dr. David Bohnhoff for all the time, effort, guidance, andfunding he put forth for this project. We would also like to thank the Biological SystemsEngineering Department at The University of Wisconsin-Madison for the use of theirequipment and facilities.

Table of Contents Page #

1. Statement of Need………………………………………………………………. 1

1.1 Post-Frame Buildings………………………………………………….. 1

1.2 Concrete Piers…………………………………………………………. 2

1.3 Concrete-To-Wood Post Connections………………………………… 8

1.4 Design Need…………………………………………………………… 10

2. Design Goal and Specifications………………………………………………… 11

3. Overview/Approach………………………………………………………..…… 12

4. Required Strength.…………………………………………….………………… 13

5. Design.……………………….…………………………………………………. 145.1 Steel Bracket Design…………………….…………………………….. 145.2 Steel Bracket-to-Concrete Connection Design………………………... 17

5.2.1 Steel Bracket-to-Concrete Connection Design One…….…… 175.2.2 Steel Bracket-to-Concrete Connection Design Two………… 31

5.3 Steel Bracket-to-Wood Connection Design…………………………………… 405.3.1 Fastener Slip-Modulus………………………………………. 415.3.2 Steel Bracket-to-Wood Fastener Designs…………………… 425.3.3 Steel Bracket-to-Wood Results and Discussion……………... 47

6. Manufacturability………………………………………………………………. 53

7. Economic Analysis….….………………………………………………….…… 55

8. Final Design………….….……………………………………………………… 56

9. Summary….………….….……………………………………………………… 62

Appendix A: Calculated Strengths………………………………………………… 63A.1 Concrete……………………………………………………………….. 63

A.1.1 Bending Strength……………………………………………..63A.1.2 Shear Strength………………………………………………..64

A.2 Wood…………………………………………………………………...66A.2.1 Bending Strength……………………………………………..66A.2.2 Shear Strength………………………………………………..67

Appendix B: Test Methods and Equipment………………………………………...68B.1 Steel Bracket-to-Concrete Tests………………………………………..68B.2 Steel Bracket-to-Wood Tests………………………………………….. 70B.3 Load-Slip Tests………………………………………………….…..… 71

Appendix C: Measured Test Results…………………………………………….…..73

References…………………………………………………………..……………….76

Index…………………………………………………………..……………………..79

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1. Statement of Need

1.1 Post-Frame Buildings

A post-frame building is a building whose primary framing system is largely or entirelycomprised of individual post-frames (Figure 1.1). A post-frame is the two-dimensionalstructure that is formed by the attachment of a wood truss (or rafters) to one or morevertical timber posts (a.k.a. columns) (NFBA, 1999b). Loads are transferred to theground through the posts that are usually embedded into the ground, and thus act as thebuilding’s foundation. In certain situations the posts may be attached to a concrete ormasonry foundation. Preservative treatment is needed when posts are embedded in theground, directly exposed to the outdoor environment, or otherwise placed in a very humidenvironment.

Figure 1.1. Example of a post-frame building

The modern post-frame building evolved from the “pole building” developed by H.Howard Doane in the 1930s (Knight, 1989). Round poles, while still used in someconstruction, were largely replaced by solid-sawn rectangular timbers in the 1960’smaking girt and truss attachment easier. During the 1980’s solid sawn lumber wasreplaced by mechanically-laminated dimension lumber. The change to mechanically-laminated posts came about due to the fact that such assemblies have a significantlyhigher bending strength to cost ratio when compared to their solid-sawn counterparts.Additionally, it is easier to fully preservative-treat dimension lumber than it is thickersolid-sawn timber. Today, most posts are still mechanically-laminated although someglue-laminated assemblies are used. Table 1 shows an approximate breakdown ofmechanically-laminated post sizes currently being produced by three majormanufacturers. Note that the majority of mechanically-laminated columns are 3-plyassemblies fabricated from nominal 2- by 6-inch dimension lumber.

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Table 1.1. Production sizes of mechanically-laminated posts

Percent of Total Mechanically-Laminated Post Production

Manufacturer 3-ply nominal

2- by 6-inch posts

3-ply nominal

2- by 8-inch post

All other post

sizes

Lester (Boor, P. 2005) 62 17 21

Morton (Kotwitz, M. 2005) 80 15 5

Wick (Sloniker, M. 2005) 85 10 5

Post-frame buildings are becoming more attractive as public interest in “green” buildingsincreases. Green buildings are structures that conserve natural resources in theirconstruction, operation, and retirement. The “greenness” of post frame buildings isprimarily due to (1) their light, structural efficient wood framework that conservesbuilding materials, and (2) their embedded post foundations that eliminate the need formore expensive concrete frost walls or grade beam foundations. The use of embeddedpost foundations also enables quick construction of the building shell, which in turnprotects interior concrete flat work from precipitation, wind, and temperature extremes.

1.2 Concrete Piers

In recent years, there has been increased interest in replacing the embedded portion ofwood posts with concrete piers. These concrete piers may be either precast or cast-in-place components. Figure 1.2 shows use of precast piers developed by Perma Column,Inc. (PCI. 2004), and Figure 1.3 shows the installation of cast-in-place piers on a researchsite (Bohnhoff et al, 2002a). Both types have their advantages and disadvantages.Construction of cast-in-place piers has been enhanced by the broad availability of columnforming tubes and the manufacture of special column footing forms which attach directlyto column forming tubes such as the one shown in Figure 1.4 (Bohnhoff et al, 2003).Column footing forms enable quick fabrication of a concrete post with a bell-shapedbottom for increased bearing capacity and uplift resistance.

Figure 1.2. Perma-Columns in use (PCI, 2005a)

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Figure 1.3. Cast-in-place concrete piers used in a post-frame building for diaphragm action studies

located in Lester Prairie, MN (Bohnhoff, 2002a)

4

Figure 1.4. Concrete column forming tube

Increased interest in concrete piers can be attributed to the following seven, largelyinterrelated reasons.

1. Durability. End users have more confidence in the long-term durability of aconcrete foundation than in a preservative-treated wood foundation. This islargely due to the poor performance of many under-treated solid–sawn posts (seeFigure 1.5). It is important to note that to date, there are no documented failuresof mechanically-laminated wood posts that have been properly treated for groundcontact with chromated copper arsenate (CCA).

Figure 1.5. Treatment penetration comparison of a solid-sawn post vs. a mechanically laminated column

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2. Reduced availability and/or higher cost of CCA-treated lumber. Numerousscientific studies have shown that wood properly treated with CCA poses nosignificant harm to humans or their environment (Burkholder, 2004; Lebow,2004; Townsend, et al, 2003; Knight, 2004). Nevertheless, CCA chemicalproducers under a continual barrage of lawsuits by individuals and groups largelyconcerned about arsenate exposures worked with the US EnvironmentalProtection Agency on a voluntary phase out of CCA. Effective December 31,2003, no wood treater or manufacturer could treat wood with CCA for mostresidential uses (http://www.epa.gov/oppad001/reregistration/cca). While postsfor agricultural and commercial buildings can still be CCA-treated, the partial banon CCA significantly reduces the amount of wood that is CCA-treated, making itmore difficult and expensive to obtain.

3. Corrosiveness and other potential dangers of CCA alternatives. Simultaneous withtheir request to phase out CCA for certain applications, chemical companiesbegan promoting already existing alternative treatments, primarily AlkalineCopper Quat (ACQ) and Acid Copper Chromate (ACC). Like CCA, thesealternative treatments rely on copper toxicity for effective protection from decayorganisms. Unlike CCA, they are not time-tested and tend to leach more copperinto the environment. They also contain other dangerous chemicals likeammonia. To combat leaching, up to three or four times the copper is used ininitial wood treatment (Burkholder, 2004; Lebow, 2004). This higher copperconcentration results in additional loss of copper, and increased galvaniccorrosion when metals less noble than copper (e.g., magnesium, zinc, iron, steel,aluminum) are driven into or brought into direct surface contact with the treatedwood (Lebow, 2004). Excessive corrosion of metal fasteners is of primaryconcern to engineers focused on structural integrity and safety of buildingoccupants (Burkholder, 2004; Knight, 2004).

4. Reduced use of preservative treated lumber. Where possible, engineers try toeliminate preservative-treated lumber because (1) it costs more than non-treatedlumber, (2) it generally requires the use of more expensive, less-corrosivefasteners, and (3) preservative wood treatments are pesticides which can makeeventual disposal of preservative-treated wood problematic. Table 1.2 gives theboard-foot costs for preservative-treated and untreated lumber and Figures 1.6 and1.7 show the graphical representations. The cost of preservative treatment alonewill drive engineers to use posts featuring treated wood spliced to untreated woodin an effort to save money for posts not requiring above ground treatment. Use ofconcrete piers in this situation eliminates the treated wood altogether, as well asthe additional assembly costs associated with joining treated and untreateddimension lumber.

5. Lumber length. As indicated by Table 1.2, the 2x6 dimensional lumber variesslightly in price (on a board-foot basis) in longer lengths, whereas the 2x8dimensional lumber becomes increasingly more expensive (on a board-foot basis)in longer lengths. Additionally, dimension lumber is not readily available inlengths longer than 20 feet. When concrete piers are used, the overall length ofthe wood post is generally shortened by four to seven feet. This means engineers

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are using shorter, less expensive lumber to obtain the same building heights, andcan also build structures with 20-foot eave heights using unspliced sidewall posts.

Table 1.2. Wholesale cost of no. 1 southern yellow pine

Cost ($) per Board FootLumber

Size

Preservative

Treatment 12 foot long 16 foot long 20 foot long

None* 0.39 0.40 0.38

0.6 pcf CCA** 0.54 0.55 0.53

#1 SYP

Nominal 2-

by 6- inch0.6 pcf ACQ*** 0.72 0.73 0.71

None* 0.57 0.62 0.60

0.6 pcf CCA** 0.72 0.77 0.75

#1 SYP

Nominal 2-by 8- inch

0.6 pcf ACQ*** 0.90 0.95 0.93*Wholesale prices from Jan. 2006 (which includes shipping and handling of the lumber to the lumberyard). (Kelly, 2006)

** Based off of Woods Run Forest Products cost of 0.6 pcf ACQ treatment per 1000 board feet.

*** Based off of Universal Forest Products cost of 0.6 pcf CCA treatment per 1000 board feet.

Figure 1.6. Cost comparison (dollars per board foot vs. board length) of non-treated vs. treated 2x6

lumber based on prices fromTable1.2

0

0.1

0.2

0.3

0.4

0.5

0.6

0.7

0.8

12' 16' 20'

Non-Treated

CCA-Treated

ACQ-Treated

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lumber based on prices from Table 1.2

6. Ease of building disassembly. Agricultural and commercial buildings have arelatively short functional design life. It is therefore beneficial to be able to easilydisassemble the building components for use in a more functional structure. Thisis much easier to accomplish when wood posts are attached to concrete piers.

7. Reuse and/or disposal of treated lumber. There are only a few alternatives for thereuse of CCA treated lumber. Additionally the reuse of materials, while now aviable option, may not be allowed in 10 to 20 years. Because of the relativelysmall quantity of lumber that would actually be reusable, there is a need forproper disposal methods. Landfills are the primary disposal alternative in theU.S. and Canada, however high costs are associated with landfill disposal. Otherways of disposing of CCA-treated lumber, such as removal of the treatment fromthe wood and then reusing it in wood composites, are under investigation(McQueen, et al. 1998). Through the use of concrete piers, the need for reusingor disposing of CCA or other chemically treated columns is virtually eliminated.

0.00

0.10

0.20

0.30

0.40

0.50

0.60

0.70

0.80

0.90

1.00

12' 16' 20'

Non-Treated

CCA-Treated

ACQ-Treated

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1.3 Concrete-To-Wood Post Connections

Despite (1) the increased use of concrete piers in post-frame building construction, and(2) the number of post-frame buildings that have been erected on concrete frost walls andgrade beam foundations, very little attention has been focused on concrete-to-wood postconnections.

Some common steel brackets used by the post-frame industry to attach wood posts tocast-in-place concrete include U-brackets and L-brackets as shown in Figures 1.8 and 1.9respectively. Relative cost and design loads for some of these steel connectors bySimpson Strong-Tie and USP are given in Table 1.3 and their pictures are shown inFigures 1.10 and 1.11.

Figure 1.8. U-bracket (Bohnhoff, 2002a) Figure 1.9. L-bracket

Table 1.3. Post-to-concrete connections

Nail Uplift (lbs)Brand Connection

No.133% 166%

Cost

*USP PA66E 1060 1060 $10.49

**Simpson PB66 1410 1410 $11.49

*USP PA66E cost is from the retail price at Menards and design loads from (USP, 2006)

**Simpson PB66 cost is from the retail price at Home Depot and design loads from (Simpson, 2006)

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Figure 1.10. USP PA66E concrete-to-post connection (USP, 2006) Figure 1.11. Simpson PB66 concrete-to-post

connection (Simpson, 2006)

Connections made with the connectors in Table 1.3 are treated as pin connections indesign because of the lack of bending strength and stiffness of (1) the steel bracket-to-concrete connection, (2) the steel bracket-to-wood post connection, and/or (3) the steelbracket itself. With concrete-to-wood post connections that lack bending strength andstiffness, the building designer must rely largely upon diaphragm action or on rigidcolumn-to-truss connections to handle horizontal forces applied to the structure. Perma-Column Inc. has the only precast concrete pier currently marketed for post-framebuilding use in the United States. The steel bracket used to connect the pier to a woodpost is an integral part of the concrete pier. For this reason, the steel bracket-to-concreteconnection has more bending stiffness than one would expect to get when using theconnectors shown in Figures 1.8, 1.9, and 1.10. Nevertheless, it is important tounderstand that the bending stiffness of Perma-Column’s steel bracket and steel bracket-to-concrete connection do not behave like the wood post or the concrete pier theyconnect. Figure 1.12 shows the bending moment versus rotation relationship for Perma-Column’s PC6300 connection (the PC6300 is the assembly used for 3-ply, 2- by 6-inchposts). This relationship does not account for the extra rotation due to slip between thebracket and wood post. For comparison purposes, the moment versus end rotation of 2-and 4-foot long, 3-ply 2-by 6-inch posts (MOE = 1.7 million lbf/in2) are also shown. It isshown from this plot that the bending stiffness of the steel bracket-to-concrete connectionplus that of the bracket itself are equivalent to a 4-foot 3-ply 2- by 6-inch post section.Since the Perma-Column connection is considerably shorter than the 4- foot board usedfor comparison it is apparent that the bending stiffness of the connection is not on thesame level as the material it is intended to replace.

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Figu

re 1.12. Bending moment to rotation comparison (PCI, 2004)

1.4 Design Need

Development of a concrete-to-wood post connection with significant bending strengthand stiffness would help concrete piers with attached wood posts behave like unjointedbeams. This would enable design engineers to either reduce overall post size, or rely lesson diaphragm action or rigid frame design for building stability. This, in turn, wouldmake concrete piers more attractive to builders, which would ultimately decreasedependence on preservative-treated lumber.

0

50000

100000

150000

200000

250000

0 0.01 0.02 0.03 0.04 0.05

2ft 3-Ply 2x6

4ft 3-ply 2x6

Perma-Column 6300

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2. Design Goal and Specifications

The goal of this project is to develop a cost effective, moment resistant connector forjoining mechanically-laminated wood posts to concrete.

For this project, we established the following design specifications:

1. The design bending strength of the connection should be at least 100% of thedesign shear & bending strength of the components it connects (i.e., the woodposts and the concrete pier/slab).

2. The average bending stiffness of the connection region should be at least 75% ofthat for the mechanically laminated post.

3. The size of the connection region cannot exceed the width and depth of themechanically-laminated post.

4. The connector shall be designed for use with mechanically-laminated postconsisting of three nominal 2- by 6-inch No.1 Southern Yellow Pine wood layers.

5. Under normal expected conditions, the connection shall have a minimum designlife of 50 years.

6. The design of the bracket-to-wood connection shall conform to the NDS forWood Construction (AF&PA, 2005) and all wood-related design strength andstiffness properties shall be calculated in accordance with NDS for WoodConstruction (AF&PA, 2005) when/where applicable.

7. The design of reinforced concrete components shall conform to all requirementsof ACI 318, and reinforced concrete design strength and stiffness properties shallbe determined in accordance with ACI 318 when/where applicable.

8. Where applicable, the design strength and stiffness of steel components shall bedetermined in accordance with AISC.

9. The design shall ensure code-conforming wood post-to-bracket connections canbe made at the job-site.

10. The total final connection and post material cost shall not exceed that for the USPPA66E retail cost of $10.49, or that for the Simpson PB66 retail cost of $11.49.

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3. Overview/Approach

From a structural design perspective, the connection consisted of three elements: a steelbracket, a steel-bracket-to-concrete connection, and a steel-bracket-to-wood connection.It was critical that this concept of three separate units was embraced because it drove theentire design and testing processes. By separating the design of this connection into threeparts, a logical examination of the load transfer through the beam and connection wasable to be completed.

Target design specification 1 states that the connection must be stronger in shear andbending than the elements that it connects (i.e., it must be stronger than the wood postand the concrete pier). If it is stronger than the elements that it connects, then engineerswill generally not need be concerned about it during building design. More specifically,if in practice, the wood post and the concrete pier meet structural design criteria, then theconnection need not be checked for adequacy.

In order to ensure that the connection is stronger than the components it connects, each ofthe three elements must have axial, shear, and bending strengths that exceed those for theconcrete pier, and those for the wood post. More in-depth analysis of three elementalstrength calculations can be found in Required Strengths (Section 4).

The steel bracket considered the most logical element with which to start (Section 5.1).The steel bracket is defined as the steel portion located above the concrete pier. Thiselement of the connection needed to be designed first as it was the foundation uponwhich the rest of the design evolved.

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4. Required Strength

First and foremost, a uniform design method for evaluating the wood, concrete, and steeldesign strengths was chosen. The two design methods considered were Allowable StressDesign (ASD) and Load Resistance Factor Design (LRFD). We chose to work withLRFD for two reasons. First, LRFD evaluates materials through ultimate strength designthat more specifically examines the materials threshold allowing for direct comparisonbetween our calculated values and test values. Second, LRFD is the primary designmethod used in industry for steel and concrete design; thus, by using LRFD all materialvalues are directly comparable and comply with industry standards.

Once LRFD was chosen, shear and bending strengths for the wood and concrete werecalculated. This was done to determine the strength criteria the connection would need tomeet in accordance with target specification No. 1. The basic strength properties of thewood from target specification No. 4 and normal weight concrete of 150 pcf, with thesame size of the 3-ply 2x6 column being 4.5 in. x 5.5 in. determined the design loads forthe connection. This allowed for choosing an appropriate kind of steel used in theconnection design. The wood strength properties were determined using Load ResistanceFactor Design equations found in the 2005 NDS (National Design Specifications forWood Construction). The concrete strength properties were determined using the ACI318-05 (American Concrete Institute) specification. The complete set of equations andsteps taken to find the strength properties for the wood and concrete can be found inAppendix A. The final determined strengths for the wood and concrete can be found inTable 4.1.

Table 4.1. Calculated strengths for wood and concrete

Shear Strength (kips) Bending Strength (kip-in)1Wood 6.24 109.17

2Concrete 1.57 39.84

1The wood used in these calculations was a #1 SYP 3-ply 2x6 column. 2

The concrete used in these calculations was normal 150 pcf, with the same size of the 3-ply 2x6 column being 4.5 in. x 5.5 in.

Once the controlling strengths of the concrete and wood materials were determined, itwas possible to begin with the connection design.

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5. Design

5.1 Steel Bracket Design

The steel bracket, or the portion of steel that was located above the steel pier, was thefoundation with which the connection design evolved. The design of the steel bracketbegins by using the required material strengths to determine the size, shape, and overalldimensions of the steel bracket.

The maximum width of the bracket was established by target specification three. Thisdesign goal established a maximum bracket width of 5.5 inches. From Equation 5.1 andthe chosen width of 5.5 inches the steel thickness needed to transfer the complete bendingmoment of the wood (109 kip-inches) through the steel was calculated.

M = fy*I/c (5.1)Where:

fy = Yield Stress of Steel= 60 ksi

I = 1/12*(b*h3)b = Thickness Needed

= (unknown)h = Width of plate

= 5.5 inchesc = Depth to the Centriod

= 2.75 in4

From the use of Equation 5.1 it was found that a 7 gauge steel plate (thickness equaling0.1875 inches) and a width of 5.5 inches had a moment capacity of 56.7 kip-inches. The56.7 kip-inches value was then multiplied by two because two plates were chosen to formthe bracket. Two plates were then chosen to form the bracket for two main reasons.First, by using two plates it is possible to symmetrically load the connection whichmaximized the material strength. Secondly, by adding additional plates themanufacturability was decreased and the cost was increased. The resulting momentcapacity for the 7 gauge steel was 113.4 kip-inches. This final value showed the steel hadthe ability to completely transfer the bending moment from the wood; thus, 7 gauge steelwas chosen for the bracket design. It should be stated that the moment capacity of theplate is found with the assumption that a buckling failure is prevented. This possible typeof failure is discussed in further detail later on.

Once the needed width and thickness of the steel plate had been determined the requiredplate length was calculated. This was done by using the material strengths found in Table4.1. First, the moment capacity needed to be transferred by the wood (109 kip-inches)was divided by shear capacity of the wood (6.24 kips). This step revealed the plate lengthto be 17.5 inches. Due to the high cost of steel in today’s market, the 17.5-inch platelength was reduced to 14 inches. With the uncertainty and variability in wood strengthproperties the reduced length of 14 inches for the steel plate would adequately suffice.

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The 14-inch plate length only required a slightly larger shear capacity of 1.56 kips in thewood to be acceptable.

The preceding steps lead to the conclusion that a bracket with a 5.5-inch width, 14-inchlength, and a 0.1875-inch thickness would have the ability to transfer the momentthrough the 3-ply 2x6 into that of the steel bracket. Next, it was necessary to determinethe placement of the bracket.

It was clearly evident that there were two main options available for the bracketplacement. These were to place the steel bracket on the outer edges of the column asdone by Perma-Column in Figure 5.1 or to place the steel bracket between the plies of the3-ply 2x6 column.

Figure 5.1. Perma-Columns in use (PCI, 2005a)

Each option for bracket placement had advantages and disadvantages that were analyzed.Advantages of the bracket placement on the outside of the columns included quickassembly and accurate placement of bolts through the use of pre-drilled holes in the steelplates. The disadvantages of outside placement include the creation of a U-shape or theneed for a wider concrete pier. If a U-shape with a thin base is created, as is done withPerma-columns, a very significant problem develops when the bracket is placed under abending load. As a U-shaped object is placed in bending it will attempt to straighten as ittransfers the applied load. This attempt to straighten crushes the concrete or buckles thesteel and ultimately renders the column useless. Figure 5.2 shows a typical failurewitnessed during Perma-column testing.

16

Figure 5.2. A typical Perma-column post failure

The second option was to place the steel plates of the bracket between the plies of the 3-ply 2x6 column which offered two significant structural benefits not seen with the firstoption. The structural benefits of placing the plates between the plies included (1) thebrackets being kept in a straight line through out the entire connection. This offered themost direct and efficient path of load transfer through the wood and concrete. (2) Theouter plies of the 3-ply 2x6 post constrain the entire area of the steel brackets. The onedisadvantage of this setup was the difficulty in quick fastener placement which resultedin an increase in the overall assembly time.

After analyzing both option’s pros and cons, the placement of the brackets between theplies of the 3-ply 2x6 column became the logical choice. This design was ultimatelychosen because the structural integrity of the column must be of primary concern and notbased solely on how easy it would be assembled.

Once the steel bracket dimensions and placement were determined, the final steps to thebracket design were to determine what variables would be examined during testing andwhat possible failure modes may occur. It was determined that the most probable failureduring testing would be the buckling of the steel plates. Analysis for this type of failurefrom using Equation 5.2 determined that the 7 gauge steel plates used would resist abending load of 115 kip-inches before reaching a buckling failure.

Mcr = (!/L)*(EIyGJ)0.5 (5.2)Where:

L = Unbraced Length= 14 in

E = Modulus of Elasticity= 29000 ksi

Iy = Moment of Inertia along the minor axis= 7.14e-3 in4 (7 gauge steel)

_ = Poisson's Ratio for steel= 0.28

G = E/2(1+_) - Shear Modulus= 11328.1 ksi

17

J = Lb3/3 - Torsional Shear Modulus

= 0.028 in4 (7 gauge steel)

A buckling failure would be unlikely to occur because the115 kip-inches bucklingstrength for the steel plates was greater than the 109 kip inches bending strength of thewood. However, since the plates were sandwiched between the 2x6 plies, someadditional lateral constraint would be exerted on the steel plates helping to resist abuckling failure. What was still unknown was if this additional constraint would allowfor the use of thinner steel. The use of thinner steel would save money; therefore, asecond type of steel (10 gauge, 0.1345 inches thick) was chosen to be another variableexamined during testing. Recalculating the buckling strength with the appropriate Iy =3.01e-3 in4 and J = 0.012 in4 indicated that the 10 gauge steel would buckle at a bendingload of 48.9 kip-inches. Testing would allow for determining if the additionalcompressive force would be significant enough to keep the thinner steel from buckling,and if so, the justification for its use. After designing the steel bracket it was possible tobegin designing the interface between the steel bracket and concrete portion of the beam.

5.2 Steel Bracket-to-Concrete Connection Design

The main objective in the steel bracket-to-concrete design was to transfer the bendingmoment from the steel bracket into the supporting rebar and concrete. The load transferfor this section was very complex as it involved the movement of the load through theconcrete to steel. Due to the complexity of this load transfer no standard procedure hasbeen developed for analyzing this interface. For this reason testing was needed toexamine this portion of the design.

5.2.1 Steel Bracket-to-Concrete Connection Design One

The steel bracket-to-concrete design process began with determining the correct lengththat effectively transferred the moment forces from the steel to the concrete. Thecalculated concrete bending strength of 39.87 kip-inches was divided by the concreteshear strength of 1.57 kips. This led to a desirable length of 25 inches. As was the priorcase with the steel bracket, the length was deemed too long due to the current steel prices;therefore, a shorter, cost acceptable embedment length of 15 inches was chosen. Thislength required a slightly larger concrete shear strength of 2.65 kips for complete loadtransfer. Furthermore, after choosing the length of 15 inches for embedment an evenshorter embedment length of 10 inches was also chosen. The shorter length allowed forfurther examination of the load transfer through the steel to the concrete and how great ofan effect the steel length had on that transfer.

As stated earlier, one of the main concerns was the possibility of the steel plate buckling.Within the concrete region such a failure mode that would likely be precipitated by theconcrete popping off the side of each plate. To combat the possible buckling failure ofsteel plates within the concrete four main modifications to the steel were discussed. First,a bend could be placed in the edge of the steel to reduce buckling potential. Second, ataper to the edge of the steel can be made. A tapered, or reduced cross-section is possiblebecause the bending moment is the steel decreases toward the end of the steel. With a

18

reduced cross-section, the steel cuts though less of the concrete making it harder to“split” off the concrete. Third, holes can be placed in the steel allowing concrete to runcontinuously across portions of the connection thus reducing the likelihood of concretesplitting. Finally, the two steel plates can be connected with rebar bracing the laterallyunsupported length of the each plate thereby reducing the buckling potential. Afterdetermining the length and keeping those four ideas in mind, several decisions weremade. Figure 5.3 illustrates a typical connection design described in the followingsection.

Figure 5.3. Example connection design

First, it was decided that all connections would include punched holes in the steelbrackets. A hole size of 1.25 inches in diameter was chosen to be placed in the plates asthis was the largest hole possible with the equipment available. Second, all plates wouldbe secured together using rebar. Four rebar segments, two at the top and two at thebottom of the plate, with a 5/8-inch spacing for the transverse rebar was chosen. Inaddition to reducing the plate buckling potential these bars help attach and transfer loadto the lengthwise rebar that would reinforce the concrete. Keeping these design elementsconstant, the effects of either a taper or bend on the edge of the steel were observed. Thedimensions of the taper and bend, shown in subsequent figures, were chosen to allow foradequate spacing between the punched holes and the edge of the steel.

After deciding on the aforementioned variables, four designs were developed: (1) a 10-inch long connection with a bent edge, (2) a 10-inch long connection with a tapered edge,(3) a 15-inch long connection with a bent edge, and (4) a 15-inch long connection with atapered edge. These designs were then constructed using the two steel thicknesses and

19

then each duplicated once to check for consistency with each design. The experimentaldesign for set one of the steel bracket-to-concrete connections is summarized in Table5.1. Note that each test specimen is denoted with a four character ID. The first characterreferences the steel thickness, being classified as either “W” wide for 7 gauge, or “T” thinfor 10-gauge steel. The second character refers to either a long (15-inch) connection as“L” or short (10-inch) connection as “S”. The third ID character refers to the taper orbend placed in the connection, with the taper as “T” and the bend as “B”. The finalcharacter in the ID refers to the batch of concrete that was used during testing. It isimportant to note that an error was made when pouring concrete, several bracket designswere covered with concrete from the same batch. The first concrete batch should havebeen poured on a single set of designs and the second batch of concrete should have beenpoured on the second set of the same designs. This should have been done to block outthe variable of concrete strength during testing. These four parts are combined to formthe final ID of each connection design for example; a thick-long-bend connection fromconcrete batch two is noted as WLB2.

Table 5.1. Experimental design for steel bracket-to-concrete connection test set one

Thick (7 gauge) Thin (10 gauge)

Long (15 in) Short (10 in) Long (15 in) Short (10 in)

WLB2 WSB2 TLB1 TSB1Bend

WLB2 WSB2 TLB1 TSB1

WLT2 WST2 TLT1 TST1Taper

WLT1 WST2 TLT2 TST1

The reason for an experimental design with such a wide range of connection designs wasthat with such a complex connection, predicting the behavior was virtually impossible.

The four connection designs constructed are shown from a side, top and isometric view inFigures 5.4 thru Figure 5.15. End views of the taper and bend type connections can beseen in Figures 5.16 and 5.17, respectively.

20

Figure 5.4. 10-inch bend connection isometric view

21

Figure 5.5. 10-inch bend connection top view Figure 5.6. 10-inch bend connection side view

22

Figure 5.7. 10-inch taper connection isometric view

23

Figure 5.8. 10-inch taper connection top view Figure 5.9. 10-inch taper connection side view

24

Figure 5.10. 15-inch bend connection isometric view

25

Figure 5.11. 15-inch bend connection top view Figure 5.12. 15-inch bend connection side view

26

Figure 5.13. 15-inch taper connection isometric view

27

Figure 5.14. 15-inch taper connection top view Figure 5.15. 15-inch taper connection side view

28

Figure 5.16. End view of the tapered connection

Figure 5.17. End view of the bend connection

All testing details, test setup, and material properties can be found in Appendix B. Thetesting results from the first set of concrete connections shown in Table 5.2 were

29

disappointing with the concrete pier failing at very low loads. It is important to note thatthe concrete failed at such low loads and not the connections. Figure 5.18 shows atypical failure from test set one and Table 5.2 shows the max load applied to the concreteconnections before failure. The values given in Table 5.3 are the max bending strength ofeach pier. This was obtained by multiplying one half the maximum applied load by the19-inch distance between the joint and the load point.

Figure 5.18. A typical failure observed during concrete testing

Table 5.2. Max load strength applied to the concrete-steel connections test set one (lbs)

Maximum Applied Loads (lbs)



841 1249 554 585Bend

1112 963 658 601

1537 1306 688 753Taper

1472 1323 1654 486

30

Table 5.3. Maximum bending strength of the concrete-steel connections test set one (kip-inches)

Maximum Bending Strength (kip-inches)



8.0 11.9 5.3 5.6Bend

10.6 9.2 6.3 5.7

14.6 12.4 6.6 7.2Taper

14.0 12.6 15.7 4.6

Due to the low load failures of the concrete it was important to determine the actualstrength of the concrete and verify that the bending moment was transferred through theconnection. A 28-day compression strength test, outlined in Appendix C, was runrevealing that the concrete did indeed possess very little compression strength. Theresults of the compression test showed that for test set one the compression strength ofthe concrete was approximately 2100 lbs. This strength was then further reduced toaccount for the fact that column testing took place after only fourteen days. It was foundthat the concrete, at the time of testing, possessed approximately 1900 lbs of compressionstrength. This strength was then used to find the actual bending strength of the concretefrom test set one. Based on these numbers the calculated bending strength for test set onewas determined to be approximately 27.3 kip-inches. Table 5.4 shows the percentage ofthe bending strength transferred, i.e. stiffness of the connections.

Table 5.4. Percentage of bending strength transferred in the connections for test set one

Percentage of Bending Strength Transferred



29.5 43.5 19.2 20.8Bend

77.3 67.1 22.9 20.9

53.6 45.5 24.0 26.2Taper

51.4 46.0 57.6 16.9

After completion of test set one it was found that the connections using the taper designoutperformed the bend design. One reason this may have occurred was because it wasfound that when removing the concrete piers from the casts, the concrete did not fullyinfiltrate the areas around the bend. These gaps could have caused a significant drop inmoment transfer through the steel to the concrete. It was also found that the longer 15-inch connections transferred significantly more of the moment when compared to theshorter 10-inch connections. This was to be expected especially considering that all

31

failures witnessed took place in the concrete. Unfortunately, a definite judgment on theoverall effectiveness of one connection when compared to a second could not be made.This was due to the fact that all failures were witnessed in the concrete; therefore, it couldbe stated that no connections failed.

5.2.2 Steel Bracket-to-Concrete Connection Design Two

Upon completion of the first set of tests it was decided a second set of steel bracket-to-concrete connections were to be designed, fabricated, and tested. Problems encounteredin the first set included the large amount, and thus high price of steel required for eachconnection. Manufacturability problems discussed in a later section were also consideredin redesigning the connection. Consequently the second test set was approached with afocus on manufacturability and cost.

The new emphasis on manufacturability and cost led to the idea that a rebar cage could beformed and welded to a shorter steel bracket Figure 5.19. The rebar cage would replacethe previous tapered and bent designs. From a cost perspective forming a rebar cagewould be relatively inexpensive in comparison to the price of the steel counterpart.

With this rebar cage idea the design of the steel bracket-to-concrete connection began bydetermining the rebar cage length. As was stated earlier, a length of 25 inches waspredicted to be the distance at which maximum shear and bending moment would besimultaneously reaching during most loadings. In test set one a length of 15 inches waschosen with cost considerations in mind; however, by using rebar it was possible to comemuch closer to the calculated 25-inch length. A length of 24 inches for the rebar cagewas selected to reduce waste. Rebar was purchased in 10-foot lengths and then cut inhalf to 5-foot lengths. Each 5-foot piece was then bent to form a U-shape with a diameterof 2.5 inches with the ends evened off leaving a length of 24 inches. This length allowedfor maximum use of a 10-foot piece of rebar and nearly satisfied the initial calculatedlength. A second length of 15 inches was chosen for an alternate connection design.This length was chosen to allow for a direct comparison between the 15-inch connectionsfrom design one.

A U-shape was chosen for all rebar brackets to increase bracket pull-out resistance. It wasalso relatively simple and straightforward to manufacture. The u-shape diameter of 2.5inches was chosen to allow for approximately 1 inch of concrete cover on the outside ofthe rebar and roughly 1.25 inches of concrete cover between the main reinforcing rebarand the brackets.

After deciding on the shape and size of the rebar cages it was decided that it would beworthwhile to try two #4 reinforcing rebar throughout the concrete pier for one set ofconnections. To mimic test set one, a concrete pier with a single #5 reinforcing rebarwould be used with a second set of connections. Thorough examination of this variationwould help to provide a clear understanding of how exactly the load moved through thesteel bracket into the rebar and concrete. Finally, 7 gauge and 10 gauge steel platethickness would be examined to be consistent with test set one.

32

It should be noted that along with the second set of designs, a pair of 7 gauge, 15-inchtapered connections from test set one were fabricated. By prefabricating this design andtesting it with second set of connections a direct comparison between each test set can bemade. The 7 gauge, 15-inch tapered connection was chosen to be included in the secondset of testing over the other designs from the first set because this design had the highestaverage load before failure during the first set of testing.

Table 5.5 below summarizes the experimental design for test set two in the same fashionthat was done with test set one. This table once again identifies the connection testspecimens for each steel-to-concrete connection with a four character ID. As in the firstset of tests the first character references the thickness of steel, being classified as either“W” wide 7 gauge, or “T” thin 10-gauge steel. The second character refers to either along, 24-inch connection, as “L” or short, 13-inch connection, as “S”. The third characterin the ID refers to the number of rebar placed in the concrete for steel reinforcement witha single #5 piece of rebar as “S”, and two #4 rebar as “D”. The final character in the IDrepresents the concrete batch with batch 1 as “1” and batch 2 as “2”. These four parts areagain combined to form the final ID of each connection design, for example; a thick-longconnection with a double rebar configuration from concrete batch 2 is noted as WLD2.Furthermore, the concrete for the second batch of connections was poured with eachbatch covering one of the two identical connections and the second batch covering thesecond of the two connections. Therefore, the concrete could not be considered avariable like it was in the first set of tests.

Table 5.5. Experimental design for steel bracket-to-concrete connection tests set two


Single Double Single Double

Batch 1 WSS1 WSD1 TSS1 TSD1Short

Batch 2 WSS2 WSD2 TSS2 TSD2

Batch 1 x WLD1 x TLD1Long

Batch 2 x WLD2 x TLD2

Figures 5.19 thru Figure 5.27 show the final connection designs and dimensions for testset two from a top, side, and isomeric view.

33

Figure 5.19. 15-inch double rebar connection isometric view

34

Figure 5.20. 15-inch double rebar connection top view Figure 5.21. 15-inch double rebar connection side view

35

Figure 5.22. 24-inch double rebar connection isometric view

36

Figure 5.23. 24-inch double rebar connection top view Figure 5.24. 24-inch double rebar connection side view

37

Figure 5.25. 15-inch single rebar connection isometric view

38

Figure 5.26. 15-inch single rebar connection top view Figure 5.27. 15-inch single rebar connection side view

39

The testing results from the second set of concrete connections shown below in Table 5.6 yieldedresults with the concrete failing at very low loads; however, the loads witnessed were higher thanwith test set one. Table 5.7 shows the max bending strength of each connection; this was obtainedby multiplying one half maximum applied load by the 19-inch distance between the joint and theload point. Once again it is important to note that the concrete failed at such low loads and not theconnections.

Table 5.6. Max load applied to the concrete-steel connections test set two (lbs)



Batch 1 894 1193 718 1396Short

Batch 2 1193 1220 946 1331

Batch 1 x 1218 x 1109Long

Batch 2 x 1469 x 1339

Table 5.7. Max bending strength of the concrete-steel connections test set two (kip-inches)



Batch 1 17.0 22.7 13.7 26.0Short

Batch 2 22.7 23.2 18.0 25.3

Batch 1 x 23.2 X 21.1Long

Batch 2 x 27.9 X 25.5

As with test set one a 28-day compression strength test, found in Appendix C, was runand revealed that the concrete once again possessed very little compression strength. Thestrength for test set two was determined to be approximately 26 kip-inches for the singlerebar design, and 54 kip-inches for the double rebar design. This bending strength wasthen compared to the actual measured bending strength to determine what percentage ofthe moment was transferred. The results showing the percentage of the calculatedbending moment transferred through the connections can be seen in Table 5.8.

40

Table 5.8. Percentage of bending strength transferred in the connections for test set two



Batch 1 65.9 43.3 52.9 49.6Short

Batch 2 82.1 41.5 65.0 45.3

Batch 1 x 44.2 x 40.2Long

Batch 2 x 49.9 x 45.6

After completion of test set two it was found that all connections transferred a significantportion of the bending moment. Unfortunately as with test set one a definite judgment onthe overall effectiveness of one connection when compared to a second could not bemade. This was due to the fact that all failures were witnessed in the concrete; therefore,no connections failed.

5.3 Steel Bracket-to-Wood Connection Design

As stated earlier, the steel bracket-to-wood connection design involved having the steelbracket ran in between the post plies. The next step for the steel bracket-to-woodconnection design would be how to securely fasten the bracket to the wood. Possiblevariables securing the column to the bracket included bolts steel piercing screws. It wasdecided to use a single fastener type being the wood-to-steel piercing screw with a _ inchdiameter and a 2 _ inch length for the entire bracket to column attachments (Figure 5.28).

Figure 5.28. Wood-to-steel piercing screw with a _ diameter and 2 _ inches length

Bolts were initially considered for use because they would provide the best way to insurethe steel bracket remained compressed between the post plies; thus, reducing buckling ofthe compression edge of the plates. Although this might have proven to be a tremendousadvantage predrilled bolt holes would have added additional fabrication time. Afterinserting the bracket between post plies, some way to very accurately drill a hole throughthe wood and the steel bracket would need to be worked out. It was decided that becauseof these logistical problems and the added cost incurred with more fabrication that usingbolts was not a feasible option. No holes would need to be predrilled and simple onsitepost construction would remain a viable option if a single type of wood-to- steel piercingscrew was used. The length of 2 _ inches was chosen for the screw’s length because itgave complete penetration through the first ply, the steel plate, and still hadapproximately one-half inch of penetration into the second ply.

41

5.3.1 Fastener Slip-Modulus

The purpose of running the load slip tests was to determine the fastener slip modulus andultimately determine the effective EI (modulus of elasticity multiplied by the moment ofinertia)of the joint region during wood testing. The slip modulus of a fastener is the ratioof the shear load transferred by the fastener, and the slip between the two componentsconnected by the fastener (interlayer slip). The data collected from the load slip tests wasused to determine the slip modulus. Once the slip modulus was determined for a singlefastener it was possible to assign a slip modulus for an entire group of fasteners. Usingthe group fastener slip modulus the rotational stiffness of the joint region was determinedand an effective EI was assigned to the joint region for use in modeling. These testresults allow full assessment of the stiffness in the steel-to-wood region.

The rotational stiffness of each joint was calculated after establishing a slip modulus forthe connection. Rotational stiffness is defined as being the ratio of the applied momentM, to the rotation of the joint region _. This is shown in Figure 5.29.

Figure 5.29. The rotation (_) between the steel bracket and wood post when a bending moment is

applied to two fastener groups spaced a distance “s” apart (PCI, 2004).

42

Equation 5.3 was used to determine the relationship between M and _.

M/_ = k*s2/2 (5.3)Where:

M = Applied bending moment_ = Joint rotation (degrees)k = Slip-modulus for a fastener or fastener groups = Spacing between fasteners or fastener groups

The applied moment M used for calculations was determined through the testing found inAppendix B. After the slip-modulus for a single fastener was quantified the slip-modulusfor the entire fastener group was determined by multiplying the number of screws in eachconnection by the slip-modulus for a single fastener. For example, if the group contained8 screws for 7-gauge steel, the slip modulus for the group would equal 8 x 21,800 lbf/inor 174,400 lbf/in.

5.3.2 Steel Bracket-to-Wood Fastener Designs

The particular (Tek 4) wood-to-steel piercing screw chosen did not have any designvalues. In lieu of the fastener design values, the lowest of the max applied loadsdetermined during the load-slip tests (explained in the following section) was used. Thelowest max load of 3294 lbf was used to establish a margin of safety similar to howLRFD establishes resistance values from the lower 5% exclusion value. In this case 83%of the tested max loads were greater than the lowest max load. The load-slip test wasconducted using two fasteners so the load of 3294 lbf was divided by two equaling 1647lbf as the max load per fastener.

The assumption that the applied load would be evenly distributed to each fastener wasneeded to properly calculate the plate length. In figure 5.29 the little boxes representfastener groups. The distance between the fastener groups’ centroid for each design were9.75, 9, and 7.75 inches. The calculated bending strength of the wood members wasapproximately 109 kip-inches which made it possible to back calculate the load placed oneach fastener within a group. This was done by dividing the 109 kip-inches by thedistance between each fastener group’s centroid. The total load on each fastener groupmay be seen below in Table 5.9.

Table 5.9. Total load on each fastener group

Fastener Group Number 1 2 3Distance Between Fastener

Groups (inches)9.75 9 7.75

Load on Each Fastener

Group (kips)11.18 12.11 14.06

After the total load to each fastener group was established, the number of fasteners ineach group was determined. After using the previously stated max load per fastener of1647 lbf lead to establishing the number of fasteners in each group to be the following:fastener groups 1, 2, and 3 having 8, 10, and 14 screws, respectively. The load per screwwas determined by dividing the load on each fastener group by the total number of

43

fasteners used. These results (shown in Table 5.10) were less than the stated max load of1647 lbf for each fastener showing that the bending strength of the wood was completelytransferred through the steel.

Table 5.10. Load applied per screw

Number of Screws 8 10 14Load per Screw (lbf) 1397.5 1211 1004

The eight screw configuration can be found in Figure 5.30. A total of eight screws, fouron each side of the column, were chosen to determine if the actual strength was higherthan anticipated resulting in the use of fewer screws and a save cost. The ten-screwconfiguration, found in Figure 5.31, was chosen not only for less resulting load perfastener, but to also help add rotational constraint near the top of the steel plate. Finally,the fourteen screw configuration, found in Figure 5.32, was chosen to find the effects ofusing several more screws than initial designs dictated to help compress the steel betweenthe wood plies in order to counteract possible steel buckling.

From the 14-inch plate length specified earlier a proper screw placement was established.This was done by finding how close a screw could be placed to the bottom of the board.The National Design Specifications for Wood Construction (NDS) recommends a screwdistance of 10 screw diameters from the bottom of the board to prevent the wood fromsplitting. Three inches or 12 screw diameters was chosen for the placement of the screwclosest to the bottom edge of wood in order to comply with this recommendation. Fromthis point the patterns were spaced so that the centroid distances would match the initialcalculations. Finally, a distance of a 1/2 inch between the screw and the top of the platewas chosen because it’s as close to the top as possible while still accounting for humanerror during drilling.

44

Figure 5.30. Dimensions and hole pattern for screw configuration 1


45


Every time additional screws were added to the connection the dimensions of the originalscrew placement would not change. This was done to ensure that each test would showthe effects of adding more screws into the configuration. An important dimension to noteis the three inches from the base of the post to the first screws. This distance was toensure that the wood would not split at the base of the column and compromise theintegrity of the wood. A distance of 13 _ inches was chosen for the topmost screwbecause very little distance was needed for the screw penetration into the edge of steel.Additionally, no screw aligned with a second screw in each screw configuration to ensurethat the wood column did not split.

After determining the screw configuration two other variables were considered. To stayconsistent with the concrete testing, the effect of varying the steel thickness between 7-gauge and 10-gauge steel was examined. The other variable chosen was the use of a lipor a no-lip design at the base of the steel found in Figures 5.33 and 5.34, respectively.

46

Knowing that the wood would want to rotate and deflect as load was applied, it wasdetermined that by placing a lip on the connection you would be able to prevent anyrotation beyond the base of the connection. For this reason, a lip and no-lip designelement was chosen as a connection variable.

Figure 5.33. Lip design Figure 5.34. No-Lip design

47

5.3.3 Steel Bracket-to-Wood Connection Results and Discussion

The steel bracket-to wood connection testing methods are outlined in Appendix B. Themaximum applied load and maximum bending moment sustained by each connection, isgiven in Tables 5.11 and 5.12, respectively. These values were obtained by multiplyingthe maximum applied load by the distance between the joint and the load point (19inches). Figures 5.35, 5.36, and 5.37 contain a plot of total applied load versus mid spandeflection for each of the 12 assembly tests. It is important to note that the displacementsgiven by the test data are an average of the relative displacements at the point loads andthat the relative displacements are not necessarily the same on each side.

Table 5.11. Maximum applied loads for wood connection tests (lbs)

7 gauge 10 gaugeColumn # and

Screw Design Lip No Lip Lip No Lip

1 4733 3234 3540 3275

2 5595 3640 3485 3036

3 5803 4252 3483 3684

Table 5.12. Maximum bending moment for wood connection tests (kip-inches)

7 gauge 10 gaugeColumn # and

Screw Design Lip No Lip Lip No Lip

1 89.9 61.4 67.3 62.2

2 106.3 69.2 66.2 57.7

3 110.3 80.8 66.2 70.0

48

0

500

1000

1500

2000

2500

3000

3500

4000

4500

5000

0 0.5 1 1.5 2 2.5 3

To

tal

Ap

pli

ed

Lo

ad

, lb

f

WL1

WN1

TL1

TN1

Figure 5.35. Total applied load verses midspan deflection for fastener design one

0

1000

2000

3000

4000

5000

6000

0 0.5 1 1.5 2 2.5 3

To

tal A

pp

lied

Lo

ad

, lb

f

WL2

WN2

TL2

TN2

Figure 5.36. Total applied load verses midspan deflection for fastener design two

49

Figure 5.37. Total applied load verses midspan deflection for fastener design three

The variables and numerical results of each test are shown in Table 5.13. Load versusslip curves from the six conducted tests are shown in Figure 5.38. At maximum load, theaverage slip was 0.13 inches. Each test was stopped after one of the steel plates separatedand fell down. Observations of the insides of the assembly after the failure revealed that100% of the screws completely sheared off. The results showed that the thicker 7-gaugesteel gave a much larger slip modulus compared to the thin 10-gauge steel. By averagingthe results from each type of steel, a slip modulus of 21800 lbf/inch and 13300 lbf/inchcan be assigned for our Tek 4 screw in 7 and 10 gauge steel, respectively.

Table 5.13. Variables and numerical results of each load-slip test

Steel

Thickness

Specific

Gravity

Slip

(in.)

Max Load

(lbs)

Ave. Max

Load (lbs)

Slip

Modulus

(lbf/in.)

Ave Slip

Modulus

(lbf/in.)

0.52 0.136 3294 11084

0.59 0.108 3874 201597 Gauge

0.68 0.073 3974

3714

34105

21800

0.52 0.195 4356 9941

0.59 0.134 3392 1323710 Gauge

0.68 0.125 4490

4079

16732

13300

50

Load vs Slip

0

1000

2000

3000

4000

5000

0 0.05 0.1 0.15 0.2

Joint Slip (in.)

Lo

ad

(lb

s.)

Test 1

Test 2

Test 3

Test 4

Test 5

Test 6

Figure 5.38. Load verses slip for the six conducted tests

The max bending moment, slip-modulus, fastener spacing, and joint rotation results foreach wood-to-steel bracket connection are shown in Table 5.14. Note that eachconnection test specimen is denoted with a three character ID. The first characterreferences the type of steel, being classified as either “W” wide for 7-gauge, or “T” thinfor 10-gauge steel. The second character refers to either a lip design as “L” or no-lipdesign as “N”. The third character refers to the screw design as “1” for design 1 with an8-screw configuration, “2” for design 2 with a 10-screw configuration, and “3” for design3 with a 14-screw configuration. These three parts are combined to form the final ID ofeach connection design for example; a thick, lip design with an 8-screw configuration(design 1) is noted as WL1.

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Table 5.14. Results determining the joint rotation for each wood-to-steel connection

Connection

Design

Max Bending

Moment

(kip-in.)

Slip

Modulus

(lbf/in.)

Fastener

Spacing

(in.)

Joint

Rotation

(degrees)

WL1 89.9 21800 10 12.1

WN1 61.4 21800 10 17.8

TL1 67.3 13300 10 9.9

TN1 62.2 13300 10 10.7

WL2 106.3 21800 9.33 8.9

WN2 69.2 21800 9.33 13.7

TL2 66.2 13300 9.33 8.7

TN2 57.7 13300 9.33 10.0

WL3 110.3 21800 7.875 6.1

WN3 80.8 21800 7.875 8.4

TL3 66.2 13300 7.875 6.2

TN3 70.0 13300 7.875 5.9

Several different failures were observed during these tests. A buckling failure found inFigure 5.39, was witnessed in all designs using 10-gauge steel. The buckling calculationsdescribed earlier indicated that the buckling failure should occur at a bending momentaround 48 kip-inches and 115 kip-inches for 10-gauge and 7-gauge steel, respectively.The actual max bending moments observed from testing are shown back in Table 5.12.

The average max bending moments incurred during testing were approximately 65 kip-inches and 86 kip-inches for the 10-gauge and 7-gauge steel, respectively. A bucklingfailure (Figure 5.39) was typical of all designs using 10-gauge steel. It is reasonable toassume that the average max bending moment of 65 kip-inches for the 10-gauge steel washigher than the calculated 48 kip-inches due to compression forces being placed on theplates while in between the boards resulting in additional buckling resistance.

Figure 5.39. A buckling failure typical of all designs using 10-gauge steel

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The average max bending moment of 86 kip-inches for the 7-gauge steel was lower thanthe calculated 115 kip-inches. The lower max bending moments were due to a shearfailure of the screws before steel buckling could occur. This type of failure (Figure 5.40)was witnessed in most designs using the 7-gauge steel. To be more specific, allconnections using the 7-gauge steel, except design 3 (14-screw configuration), shearedthe screws at the top of the connection. The shearing of the screws happened at a maxapplied load between 3200 and 4250 lbs. It should be noted that design 3 (14-screwconfiguration) resulted in a buckling failure for the 7-gauge steel, only this failurehappened at a higher bending moment of 110 kip-inches which was close to thecalculated value of 115 kip-inches. Since the buckling failure consistently occurredaround the same load regardless of the amount of screws neither the screw configurationnor the amount of screws affected the compression strength.

Figure 5.40. A shear failure that was typical of all designs using 7-gauge steel except for design 3 (14-

screw configuration).

This shear failure witnessed for designs 1 and 2 in the 7-gauge steel was not anticipatedat the loads witnessed. It was expected at a higher load from what the results of the load-slip tests in Table 5.13 reported. The load-slip test suggested that a single set of screws,that being one screw on each side, would fail in shear between 3200 and 4500 lbs. Thefact that this type of failure was witnessed at similar loads for connections using multiplesets of screws suggests that during testing a single set of screws may have absorbed themajority of the load and caused a failure.

Finally, a difference was also noted in the load capacity of the lip versus no-lip designs(Table 5.12). The connections with a lip design held significantly more load than the no-lip design. This can be attributed to the fact that as the load was applied to theconnection the wood became flush with the lip resulting in load being transferred to theconnection. As load was transferred to the connection via the lip, a lower load wasexperienced in the fasteners. Thus, a higher load was handled by the connections with alip. Other differences noticed with the lip and no-lip designs regarded the rotation of thejoint region. Table 5.14 showed that lip designs were allowed to rotate an average of 8.7degrees while the no-lip designs rotated an average of 11.1 degrees. This showed that thelipped design element did indeed affect the connection by resisting joint rotation.

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6. Manufacturability

Manufacturability is directly related to the cost, and hence overall feasibility of theconnection.

During construction it is desirable to limit the number of different machines, as thisreduces initial capital investment in production and can result in more efficientproduction. The Scotchman 6509-24M iron worker shown in Figure 6.1 was used for allsteel plate/sheet work which included shearing, punching, and bending. The ironworkeralso had attached scales and braces to ensure exact placement of the steel, allowing allwork to be replicated accurately. Other equipment used included a plasma torch, wirefeed welder, belt grinder, table saw, and rebar bender.

Figure 6.1. Scotchman 6509-24M used for ironworking (Scotchman, 2006)

Throughout construction of the first series test specimens iron worker performed quitewell for punching the appropriate holes and shearing the initial rectangular pieces.However, after the initial rectangular pieces had been constructed the iron worker did nothave the ability to bend and shear the connection designs to the level of detail we needed.Finishing the first set of connections required special design templates to be constructed

54

to correctly shear the taper in the steel plates. These templates were made to ensureidentical cuts in the faceplates were made every time. This additional equipment severelyaffected the total time spent in construction for the first set of connections and brought alevel of complexity not initially anticipated.

As mentioned earlier, the production of the first series of test specimens resulted in therealization for specialized equipment and a reduction in production time. A moresimplistic approach to manufacturing was taken and the rebar cage approach wasdeveloped. By using rebar in the second series of test specimens the need for complexbending or shearing was eliminated. During the production of the second series of testspecimens it was found that bending the rebar to the appropriate U-shape found in Figure6.2 was very easy, quick, and accurate when compared to the bending and shearing of thesteel plates done to manufacture the first set of connections.

Figure 6.2. Rebar bracket

Both sets of connections required the construction of a faceplate used to join the steelplates and form the steel bracket. The production of faceplates turned out to be acumbersome part of production. This design aspect was unavoidable because thefaceplate needed to be a single unit in order to properly act as the joining agent for eachof the steel plates.

Overall, the test set two rebar design had a much higher level of manufacturability thantest set one. The rebar design required fewer steps from the beginning to end along withless time needed for manufacturing.

55

7. Economic Analysis

The cost analysis of this project took a look at simply the material cost associated withconstructing the final connection design. Labor and manufacturing costs were notanalyzed due to the lack of substantial data. It was also important to note that all costslisted were not wholesale and would be reduced if the connection were to bemanufactured on a large scale basis.

The main material associated with construction of this connection was steel. For the finalconnection it was decided that 7-gauge steel would be used. The final connection designused a total of 176 in2 (16in x 5.5in x 2 plates) which was translated to 4.583 lbs afterusing a conversion factor of .0521 lbs/in2. After multiplying the weight of the connectionby the price of steel per pound, $0.50418 per lb, a total steel cost of $4.623 $ perconnection was established. This cost was then added to that of the rebar. The rebar costwas found by multiplying the total 124-inch length of rebar (54 in x 2 brackets + 4 in x 4crossbars) by the $0.0289 per inch cost for the rebar establishing a total cost of $3.585per connection. Summing the cost of steel and rebar gave a total cost of $8.209 perconnection in material costs.

Aside from material costs the expense of coating the connection needed to be examined.Although galvanization was not required because the wood did not need to be treated; itwas still a good idea to consider coating the connection to resist rusting from possiblemoisture conditions the connection may encounter while in use. After consulting localgalvanizing shops the typical cost for galvanizing an object based on the weight of theobject generally ran about $0.35 per lb. With a connection weight of 4.583 lbs anestimated cost of $1.60 needed to be added to the total connection cost.

After combining both material costs and the cost of a galvanized coating a total cost of$9.90 per connection was assigned to the final connection design.

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8. Final Design

A final connection design was assembled using the information gathered from the threesets of tests and calculations. Since no concrete-to-steel connections experienced failureit was not possible to accurately determine which connection design would provide thebest results. That being said, most likely the most efficient design would be the 24-inchdouble rebar design. The rebar design was chosen over the steel to help reduce cost asrebar is a lot less expensive than steel. The 24-inch bracket would also allow formaximum moment transfer to the concrete because of the length of the rebar, while stillallowing for continuity throughout the concrete. Seven gauge steel would also be used toform the plates because the thicker steel was necessary to prevent a buckling failure. Thefinal connection also includes the lip design element explored in testing. Use of the lipdesign would help to increase the rotational stiffness of the connection. The fastenerconfiguration of the final design comprises of a single 1/2" bolt placed at the top andbottom of the steel bracket. On top of the bolts 4 steel piercing screws, (two on eachside) should be used at the top of the connection. The final fastener configurationincluded bolts because testing showed that the steel would buckle at an unacceptably lowload regardless of the type of steel used. By using a 1/2" bolt at the top and base of theconnection the steel would have additional compression forces and better resist abuckling failure that the screws and column plies could not. Screws are used at the top ofthe connection to reduce stress concentrations around the bolt. As the screws help absorbload and stress they are ultimately helping to keep the wood from splitting along the boltline. No screws are used at the base of the connection since the lip at base of theconnection would help absorb load that the bottom bolt would have otherwise beenforced to take. It is important to note that because the final design incorporates a bolt, a1/2" steel piercing drill bit would be needed. By using a drill bit, the bolt holes can beplaced in the wood and steel at the same time ensuring proper alignment. These designelements discussed will be put together to form the final connection design. The finaldesign and dimensions can be seen in Figures 8.1, 8.2, and 8.3.

57

Figure 8.1. Side view dimensions of the final design

58

Figure 8.2. Top view dimensions of final design

59

Figure 8.3. Final Connection Design

60

After completing design, fabrication, and testing it is important to reassess the originaldesign specifications and make a determination as far as achievement of originalobjectives. The following section restates each design specification and discusses theachievement or shortcoming of each initial goal.

1. The design bending strength of the connection should be at least 100% of the designbending strength of the components it connects (i.e., the wood columns and theconcrete pier/slab). The wood-to-steel connection testing resulted in either a bucklingfailure or shear failure of the screws; therefore, not allowing us to completelyevaluate the actual bending strength of the steel connection. To better assess theactual bending strength of the wood-to-steel connection a better fastener pattern mustbe designed to prevent buckling or fastener shearing. The concrete-to-steelconnection testing resulted in a bending failure of the concrete at the end of the steelembedded in the concrete; thus, indicating that our connection’s bending strength washigher than that of the concrete pier.

2. The average bending stiffness of the connection region should be at least 75% of thatfor the mechanically laminated column. Analysis of the connection joint regionshows that for the fasteners used a stiffness of at least 100 % was determined for theconnections using 7-gauge steel meeting our specification, however the connectionusing 10-gauge steel had an average stiffness of about 65% that of the wood usedfalling short of the required stiffness.

3. The size of the connection region cannot exceed the width and depth of themechanically-laminated column. All designs constructed used a 5.5 x 4.5 inchfaceplate which served to set the dimensions and keep the total connection andconcrete column width equal to that of the mechanically-laminated columns.

4. The connector shall be designed for use with mechanically-laminated columnconsisting of three nominal 2- by 6-inch No.1 Southern Yellow Pine wood layers.Only 2- by 6-inch No.1 Southern Yellow Pine three-ply columns were considered indesign and all connections used in testing conform to this design specification.

5. Under normal expected conditions, the connection shall have a minimum design lifeof 50 years. All components used that may come into contact with the ground and ormoist conditions are comprised of concrete and steel. While the steel connectionswill need to be galvanized to ensure a longer life, all materials used have a normal lifethat far exceeds the 50 year design specification.

6. The design of the bracket-to-wood connection shall conform to the NDS for WoodConstruction (AF&PA, 2005) and all wood-related design strength and stiffnessproperties shall be calculated in accordance with NDS for Wood Construction(AF&PA, 2005) when/where applicable. Wood-related design strength and stiffnessproperties conforming to this specification can be found in Appendix B.

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7. The design of reinforced concrete components shall conform to all requirements ofACI 318, and reinforced concrete design strength and stiffness properties shall bedetermined in accordance with ACI 318 when/where applicable. Concrete-relateddesign strength and stiffness properties conforming to this specification can be foundin Appendix A.

8. Where applicable, the design strength and stiffness of steel components shall bedetermined in accordance with AISC. With such intricate internal interactionsbetween the steel embedded into the concrete, design values must be solelydetermined from laboratory tests and not from specific equations outlined inaccordance with AISC (PCI, 2004). Therefore, the rotational stiffness of the steelcomponents was determined from slip tests in accordance with ASTM 1761.

9. The design shall ensure code-conforming wood column-to-bracket connections canbe made at the job-site. For quality assurance purposes it would be recommended topre-manufacture the post. That being said, the steel connection has been fabricated sothat it is possible for the composite post to be constructed on site. However, foronsite construction a concrete form would be needed to ensure proper column sizeand allow for rebar placement.

10. The total final connection and column design cost shall not exceed that for the USPPA66E cost of $10.49, or that for the Simpson PB66 cost of $11.49.The total cost ofeach connection was determined to be $9.90. This total materials cost meets thedesign specification.

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9. Summary

The vast majority of agricultural structures are post-frame buildings. The posts in themajority of these buildings are embedded in the ground where they are a main element ofthe building’s foundation. To minimize decay, embedded wood posts must be treatedwith some type of wood preservative. However, the use of preservative-treated wood issomewhat problematic, and raises environmental concerns when the chemicals used totreat the wood columns leaches into the ground. On top of the chemical leaching issues,disposal of the columns after the useful life of the building is a concern; because of this itis advantageous to use something other than the standard embedded wood post. Onesuch alternative is to replace the embedded portion of a wood post (i.e., the portion thattypically needs to be preservative-treated) with a concrete pier. The goal of this researchproject was to develop the connection needed to adequately attach a concrete pier to awood post. More specifically, the goal was to design a connection that would have abending strength and stiffness near that of the components it was connecting.

The approach to this project involved breaking the connection into three main elements:(1) a steel bracket portion, (2) a steel bracket-to-concrete pier connection, and (3) a steelbracket-to-wood post connection. After the designing and constructing several types ofconnections, tests produced mixed results. All failures observed for the first and secondround of steel bracket to concrete testing were concrete failures at the end of theconnection region. This was a desirable result at it showed that it was the concrete thatfailed and not the connection however, the concrete failures all occurred at a much lowerload than anticipated which made it impossible to make definite distinctions betweenconnection designs. The wood-to-steel portion of testing revealed that use of a lip torestrain the wood plies increased the rotational stiffness of the beam. This increasedstiffness keeps the beam from rotating large amounts adding to the usefulness of theconnection, but more importantly it allows the connection to withstand higher load

Using the test results the final design proposal was developed for the attachment of a 3-layer, mechanically laminated post fabricated from nominal 2-by 6-inch lumber to a 4.5-by 5.5-inch concrete pier. The final proposed design consists of two 14-inch long by 5.5-inch wide plates that are 0.1875 inches thick. These plates are sandwiched between thelumber plies, and held in place with two _” bolts and four _” by 2 _” steel piercingscrews. The plates extend 2 inches into the concrete pier. Two rebars, each with a 180degree bend, and a formed length of 24 inches are welded to the steel plates. The two180-degree rebars are joined by transverse rebar welded on to form a location for thelongitudinal rebar that will be used as reinforcement in the concrete pier. Use of thisconnection should help to eliminate the problems created by using chemically treatedposts while still maintaining post frame buildings simplicity and structural efficiency.

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Appendix A: Calculated Strengths

A.1 ConcreteA.1.1 Bending Strength

Before connection design and construction could begin, the bending and shear strengthof the concrete members needed to be known. This was needed to determine theappropriate type of steel and the correct dimensions that can be used in the steel bracketdesign. The expected bending strength of a singly reinforced concrete beam wascalculated using Equation A.1. This calculation was run four times, looking at the effectof using different sized reinforcing rebar. The number 3, 4, 5, and 6 rebar were examinedfor possible use. All bending strength equations and methods were in accordance withACI 318-05 (Nawy, 2005).

Mn = (_1*fc'*b*a*(d-a/2)) (A.1)Where:

_1 = 0.85 Note: _1 = 0.85 for fc' less than or equal to 4,000 psi, 0.65 for fc'greater than or equal to 8,000 psi, and a linear interpolation isneeded to obtain _1 between 4,000 and 8,000 psi

fc' = Normal Concrete Compression Strength= 4 ksi

b = Overall Concrete Width= 4.5 inches

a = depth of equivalent rectangular stress block= (As'*fy)/( _1*fc'*b)As' = Area of steel rebar

= 0.11 in2 (#3 rebar)= 0.22 in2 (#4 rebar)= 0.31 in2 (#5 rebar)= 0.44 in2 (#6 rebar)

fy = Yield Stress of Steel= 60 ksi

d = Depth of Steel= 2.75 inches

The bending strength in kip-inches of a concrete member with each type of steel rebarcan be seen in Table A.1

Table A.1. Calculated bending strength of concrete with a single rebar configuration (kip-inches)

No. 3 Rebar No. 4 Rebar No. 5 Rebar No. 6 Rebar

Calculated

Bending Moment16.73 28.29 39.84 49.82

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The bending strength for a doubly rebar reinforced configuration is calculated usingEquation A.2

Mn = Cc*(h/2-a/2)+Cs*(h/2-d')+Ts*(d-h/2) (A.2)

Where:

Cc = Concrete Compression Strength

= _1*fc'*b*a

h = height

= 5.5 inches

Cs = Compression Steel Compression Strength

= As'*fy

d = distance to bottom steel

= 4 inches

d' = distance to top steel

= 1.5 inches

Ts = Tension Steel Compression Strength

= As*fy

The final bending strength in kip-inches of a concrete member with two number foursteel rebar was calculated to be 48.88 kip-inches.

A.1.2 Shear Strength

The nominal concrete shear strength, Vn, equals the concrete shear strength, Vc, and theshear reinforcement shear strength, Vs, forming the equation Vn= Vc+Vs. No shearreinforcement was placed in the concrete piers resulting in Vn= Vc. Note that theconcrete pier depth is 5.5 inches equaling an overall depth less than 10 inches; therefore,according to ACI 318 Section 11.5.5.1 Vn= Vc and does not have to be reduced to Vn=Vc/2. The approximate shear strength of each concrete pier when subjected to shear andflexure only can be calculated using equation A.3. All shear strength equations andmethods used are in accordance to ACI 318-05. (Nawy, 2005), (PCI, 2004)

65

Vc = _*2*(fc')1/2*b*d (A.3)

Where:_ = 1.0 Note: _ = 1.0 is factor dependent on the type of concrete, 1.0 for

normal-weight concrete, 0.85 for sand-lightweight concrete, and0.75 for all lightweight concrete.

fc' = Normal Concrete Compression Strength = 4 ksi

b = Overall Concrete Width = 4.5 inches

d = Overall Concrete Depth = 5.5 inches

The calculated design shear strength for both concrete pier designs equals 1.57 kipsbecause there is no shear reinforcement and no change in dimensions resulting in theshear strength based solely on concrete compression strength.

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A.2 Wood

A.2.1 Bending Strength

Before construction and testing the approximate bending and shear strength of the woodmembers needed to be known in order to choose the appropriate type of steel and thecorrect dimensions. In accordance to NDS 2005, the expected bending strength for woodis calculated using the following Equation A.4

M = Fb’* S (A.4)

Where:

Fb' = Fb CM Ct CLCF Cfu Ci Cr Kf !b "

Fb = Bending Strength of #1 Southern Pine= 1.65 kips

CM = Wet Service FactorCt = Temperature FactorCL = Beam Stability FactorCF = Size FactorCfu = Flat Use FactorCi = Incising FactorCr = Load Sharing Factor due to the use of three 2x6 wood members

=1.35Cc = Curvature Factor

All values used unless otherwise noted can be taken as 1.

Kf = Format conversion Factor= 2.16/ !

!b = Resistance Factor

= 0.85" = Time Effect Factor

= 1S = Section Modulus

= 22.69 in3

Thus the final bending strength of an average 3 ply 2x6 wood member is calculated to be109.17 kip-inches.

67

A.2.2 Shear Strength

The approximate shear strength of each wood column can be calculated using EquationA.5

Vu = Fv' *A / 1.5 (A.5)Where:

l = Time Effect FactorFv = Resistance Factor for ShearFv' = Fv CM Ct Ci Kf !v "

Fv = Shear Strength Parallel to Grain= 0.175 kips

CM = Wet Service FactorCt = Temperature FactorCi = Incising Factor

All values used unless otherwise noted can be taken as 1.

Kf = Format conversion Factor= 2.16/ !

!bv = Resistance Factor

= 0.75" = Time Effect Factor

= 1A = Area = Width * Depth

Width = 4.5 inDepth = 5.5 in

= 24.75 in2

Using Equation A.5 the final shear strength of an average 3 ply 2x6 wood member iscalculated to be 6.24 kips.

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Appendix B: Test Methods and Equipment

B.1 Steel Bracket-to-Concrete Tests

Two sets of concrete tests were conducted to examine the performance of severaldifferent steel bracket-to-concrete connections. Both of these sets of tests had one similargoal in mind, to determine the bending moment versus rotation relationships for theconnections. Test set one examined the relative strength of (1) brackets using bend ortaper design for the edge of the steel embedded in the concrete (2) brackets of either a 10-gauge or 7-gauge steel thickness (3) brackets with an embedment length into the concreteof 10 or 15 inches. Test set two was conducted to examine the relative strength ofconnections with a rebar cage configuration using (1) varying embedment lengths of 15and 24 inches (2) 10-gauge or 7-gauge steel bracket thickness, (3) single or doubleconcrete reinforcing rebar design. All the brackets were designed so that two identicalplates could be welded to the base forming a single bracket, and adequate concrete coverof the steel is provided on all sides.

Concrete-to-steel connection tests were performed 14 days after initial pouring instead ofthe normal 28-day time span specified by (ASTM C39) due to time constraints. Theconcrete-to-steel connections were loaded in bending to determine the flexural strength ofeach steel bracket-to-concrete column connection. The 1/3-point loading arrangementused is shown in figure B.1. The interface between the concrete and steel bracket locatedat the center of the two loads so it located in the constant moment region that is created.

Figure B.1. Two point loading set-up for testing connections bending strength

69

Figure B.2. Tinius Olson Universal Compression-Tension Testing Machine

The Tinius Olson Universal Compression-Tension Testing Machine shown in Figure B.2was used for testing. The load head rate was fixed at 0.2 inches per minute to produce afailure between 5 and 10 minutes after initial loading in accordance with ASTMstandards. A Campbell Scientific CR23X data logger was used to record applied load,load-head movement, and displacements at _-second intervals.

To isolate the failure to the concrete portion of beam a 5-foot steel tube that can be seenin Figure B.3 was attached to the connection where the wood member would normally beplaced. Four 1-inch bolts were used to connect the steel tubing to the steel brackets withouter plates bolted to the outside of the steel bracket connection to prevent buckling.Figure B.4 shows the four 1-inch bolts used to join the connection to the steel tubing.

70

Figure B.3. Steel tubing used for testing

Figure B. 4. Bolts attaching steel connection to steel tube

B.2 Steel Bracket-to-Wood Tests

A series of tests were conducted to examine the performance of different steel-bracket towood connections. These tests were conducted to determine the relative strength ofconnections using lipped and non-lipped designs and brackets consisting of 10-gauge or7- gauge steel. There was also a desire to determine the relative strengths of connectionsusing differing numbers (8,10,14) of _ -by 2 _-inch steel piercing screws.

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Twelve wood-to-steel connections were loaded in bending to determine the flexuralstrength of each steel bracket-to-concrete column connection. The 1/3-point loadingarrangement shown earlier in Figure B.1 for concrete testing was also used for woodtesting. The interface between the wood and steel bracket is once again located at thecenter of the two loads so it located in the constant moment region that is created.

The Tinius Olson Universal Compression-Tension Testing Machine shown in Figure B.2was used for testing. The load head rate was fixed at 0.4 inches per minute. A CampbellScientific CR23X data logger was used to record applied load, load-head movement, anddisplacements at _-second intervals. The same 5-foot steel tube that can be seen inFigure B.3 was used to isolate the failure to wood section. The tubing was again attachedto the connection with four 1-inch bolts to connect the steel tubing to the connection withouter plates bolted to the outside of the steel connection to prevent buckling.

B.3 Load-Slip Tests

The load-slip methods for testing were done similar to (Bohnhoff, 2002b) and inaccordance to ASTM E 290-97a. Six wood-to-steel load-slip tests were assembled fromthree 5-foot No. 2 SPF 2x6 members. The three different lumber specific gravity valueswere 0.52, 0.59, and 0.68, respectively (based on weight and volume at a moisturecontent of approximately 10%). Each test had a total of two screws and had a layoutsimilar to that of the wood to steel connection that was used in our wood column to steeltests. The two variables in this test were the wood specific gravity and the thickness inthe steel. Two thicknesses of steel, 7-gauge and 10-gauge were used with each specificgravity value lumber.

LVDT’s located on each end of the connection were used to measure the displacementbetween the top (non-stressed part of the wood) of the outside plies and the bottom of thesteel connection (or unstressed portion of the steel connection) as seen in Figure B.5.The load was applied parallel-to-grain at a rate of 0.1 inches/minute. Displacements andapplied loads were recorded at 1-second intervals

72

Figure B.5. Load-slip test set up for determining fastener slip modulus

73

Appendix C: Measured Test Results

C.1 Concrete Compression Strength

The compression strength of the 6-by 12-in. (152x305 mm) standard cylindrical testspecimen was tested 28 days after the initial pour in accordance with ASTM C 39, madeand cured in accordance with ASTM C 192, and capped in accordance to ASTM C 617-98. The compression strength can be seen in Tables C.1 & C.2.

Table C.1. Set 1 concrete cylinder compression strengths after 28 days (psi)

Concrete Cylinder # Compression Strength Average

2328Batch 1

22312280

2069Batch 2

20142042

Table C.2. Set 2 concrete cylinder compression strengths after 28 days (psi)


2076Batch 1

19692022

2066Batch 2

23122189

The cylinders were tested after 28 days while our concrete pier bending tests wereconducted after 14 days; therefore, an equation is needed to quantitatively estimate thecompression strength of the concrete after 14 days. Equation C.1 specified by the ACI209 Committee in the 2005 version of the ACI Manual of Concrete Practice estimatingthe compression strength at any given time.

(fc')t = 28)'( cfta

t

!+(C.1)

Where:a = constant

= 4.0# = constant

= 0.85t = age in days of concrete(fc’)28 = 28-day compression strength of concrete

74

Using the appropriate constants for the type of curing (moist or steam curing) in our casemoist and the type of cement used (Type I or Type III) in our case Type I, thus EquationC.2 is developed:

(fc')t = !"

#$%

&

+ )85.04()28('

t

tf c (C.2)

The resulting 14-day compression strengths using Equation C.2 are shown in Tables C.3& C.4.

Table C.3. Set 1 concrete cylinder compression strengths estimation after 14 days (psi)


2050Batch 1

19642007

1822Batch 2

17731798

Table C.4. Set 2 Concrete cylinder compression strengths estimation after 14 days (psi)


1828Batch 1

17331780

1819Batch 2

20361920

The actual concrete bending strength using the resulting 14-day compression strengthestimation for test set one is shown in Table C.5 and the test set two estimation for singleand double rebar configurations are shown in Tables C.6 and C.7, respectively.

Table C.5. Set one bending strengths using 14-day compression strengths (kip-inches)

Concrete Cylinder # Bending Strength Average

29.1Batch 1

28.128.6

26.3Batch 2

25.625.95

75

Table C.6. Bending strengths using 14-day compression strengths for single rebar (kip-inches)


26.4Batch 1

25.125.8

26.3Batch 2

28.927.6

Table C.7- Bending strengths using 14-day compression strengths for double rebar (kip-inches)


53.6Batch 1

51.152.4

53.4Batch 2

58.455.9

76

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Environment. 309: 173-185.Universal Forest Products, 2006. Personal Communication. 3 January.USP. 2006. USP Anchor to Connector Guide 2005. United Steel Products Company.

Available at: http://www.uspconnectors.com/pdfs/2005acguide.pdfAccessed 1/23/2006.

Walker, J. and F. Woeste. 1992. Post-Frame Building Design. 1st ed. St. Joseph, MI:American Society of Agricultural Engineers.

Woods Run Forest Products. 2006. Personal Communication. 3 January.

2005 National Design Specification® for Wood Construction. 2005 American WoodCouncil and American Forest & Paper Association

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INDEX

List of Figures Page #

Figure 1.1. Example of a post-frame building……………………………………………………………….. 1

Figure 1.2. Perma-Columns in use (PCI, 2005a)…………………………………………………………….. 2

Figure 1.3. Cast-in-place concrete piers used in a post-frame building for diaphragm action studies,

Located in Lester Prairie, MN (Bohnhoff, 2002a)………………………………………………. 3

Figure 1.4. Concrete column formingtube…………………………………………………………………… 4

Figure 1.5. Treatment penetration comparison of a solid-sawn post vs. a mechanically laminated column… 4Figure 1.6. Cost comparison (dollars per board foot vs. board length) of non-treated vs. treated 2x6

lumber based on prices fromTable1.2……………………………………………………………. 6


lumber based on prices from Table 1.2…………………………………………………………… 7

Figure 1.8. U-bracket (Bohnhoff, 2002a)…………………………………………………………………….. 8

Figure 1.9. L-bracket…………………………………………………………………………………………. 8

Figure 1.10. USP PA66E concrete-to-post connection (USP, 2006)………………………………………….. 9

Figure 1.11. Simpson PB66 concrete-to-post connection (Simpson, 2006)…………………………………… 9

Figure 1.12. Bending moment to rotation comparison (PCI, 2004)…………………………………………… 10

Figure 5.1. Perma-Columns in use (PCI, 2005a)……………………………………………………………... 15

Figure 5.2. A typical Perma-column post failure…………………………………………………………….. 16Figure 5.3. Example connection design………………………………………………………………………. 18

Figure 5.4. 10-inch bend connection isometric view………………………………………………………..... 20

Figure 5.5. 10-inch bend connection top view………………………………………………………………… 21

Figure 5.6. 10-inch bend connection side view……………………………………………………………….. 21

Figure 5.7. 10-inch taper connection isometric view ……………………………………………………….... 22

Figure 5.8. 10-inch taper connection top view ……………………………………………………………….. 23

Figure 5.9. 10-inch taper connection side view……………………………………………………………….. 23

Figure 5.10. 15-inch bend connection isometric view …………………………………………………………. 24

Figure 5.11. 15-inch bend connection top view ……………………………………………………………….. 25

Figure 5.12. 15-inch bend connection side view ………………………………………………………………. 25

Figure 5.13. 15-inch taper connection isometric view ………………………………………………………… 26

Figure 5.14. 15-inch taper connection top view ……………………………………………………………….. 27Figure 5.15. 15-inch taper connection side view……………………………………………………………….. 27

Figure 5.16. End view of the tapered connection………………………………………………………………. 28

Figure 5.17. End view of the bend connection…………………………………………………………………. 28

Figure 5.18. A typical failure observed during concrete testing………………………………………………. 29

Figure 5.19. 15-inch double rebar connection isometric view ……………………………………………….. 33

Figure 5.20. 15-inch double rebar connection top view ………………………………………………………. 34

Figure 5.21. 15-inch double rebar connection side view ……………………………………………………… 34

Figure 5.22. 24-inch double rebar connection isometric view ………………………………………………… 35

Figure 5.23. 24-inch double rebar connection top view .………………………………………………………. 36

Figure 5.24. 24-inch double rebar connection side view……………………………………………………….. 36

Figure 5.25. 15-inch single rebar connection isometric view ………………………………………………….. 37Figure 5.26. 15-inch single rebar connection top view ……………………………………………………….... 38

Figure 5.27. 15-inch single rebar connection side view ……………………………………………………….. 38

Figure 5.28. Wood-to-steel piercing screw with a _ diameter and 2 _ inches length………………………..... 40

Figure 5.29. The rotation (_) between the steel bracket and wood post when a bending moment is

applied to two fastener groups spaced a distance “s” apart (PCI, 2004)…………………………… 41

Figure 5.30. Dimensions and hole pattern for screw configuration 1…………………………………………… 44

Figure 5.31. Dimensions and hole pattern for screw configuration 2…………………………………………... 44

Figure 5.32. Dimensions and hole pattern for screw configuration 3…………………………………………… 45

Figure 5.33. Lip design……………………………………………………………………………………….…. 46

Figure 5.34. No-Lip design…………………………………………………………………………………….... 46

Figure 5.35. Total applied load verses midspan deflection for fastener design one…………………………….. 48

Figure 5.36. Total applied load verses midspan deflection for fastener design two……………………………... 48

80

Figure 5.37. Total applied load verses midspan deflection for fastener design three……………………………. 49

Figure 5.38. Load verses slip for the six conducted tests………………………………………………………… 50

Figure 5.39. A buckling failure typical of all designs using 10-gauge steel……………………………………… 51

Figure 5.40. A shear failure that was typical of all designs using 7-gauge steel except for design 3 (14-

screw configuration)……………………………………………………………………………….. 52Figure 6.1. Scotchman 6509-24M used for ironworking (Scotchman, 2006)………………………………….. 53

Figure 6.2. Rebar bracket……………………………………………………………………………………….. 54

Figure 8.1. Side view dimensions of the final design…………………………………………………………... 57

Figure 8.2. Top view dimensions of final design……………………………………………………………….. 58

Figure 8.3. Final Connection Design………………………………………………………………………….… 59

Figure B.1. Two point loading set-up for testing connections bending strength………………………………… 68

Figure B.2. Tinius Olson Universal Compression-Tension Testing Machine…………………………………… 69

Figure B.3. Steel tubing used for testing………………………………………………………………………… 70

Figure B.4. Bolts attaching steel connection to steel tube………………………………………………………. 70

Figure B.5. Load-slip test set up for determining fastener slip modulus………………………………………… 72

81

List of Tables Page #

Table 1.1. Production sizes of mechanically-laminated posts………………………………………………. 2

Table 1.2. Wholesale cost of no. 1 southern yellow pine………………………………………………......... 6

Table 1.3. Post-to-concrete connections………………………………………………................................... 8

Table 4.1. Calculated strengths for wood and concrete………………………………………………............ 13

Table 5.1. Experimental design for steel bracket-to-concrete connection test set one………………………. 19

Table 5.2. Max load strength applied to the concrete-steel connections test set one (lbs)…………………… 29

Table 5.3. Maximum bending strength of the concrete-steel connections test set one (kip-inches)…………. 30

Table 5.4. Percentage of bending strength transferred in the connections for test set one…………………… 30

Table 5.5. Experimental design for steel bracket-to-concrete connection tests set two……………………… 32

Table 5.6. Max load applied to the concrete-steel connections test set two (lbs)……………………………. 39

Table 5.7. Max bending strength of the concrete-steel connections test set two (kip-inches)……………….. 39

Table 5.8. Percentage of bending strength transferred in the connections for test set two…………………... 40Table 5.9. Total load on each fastener group………………………………………………............................. 42

Table 5.10. Load applied per screw………………………………………………........................................... 43

Table 5.11. Maximum applied loads for wood connection tests (lbs)………………………………………... 47

Table 5.12. Maximum bending moment for wood connection tests (kip-inches)……………………………. 47

Table 5.13. Variables and numerical results of each load-slip test…………………………………………... 49

Table 5.14. Results determining the joint rotation for each wood-to-steel connection……………………… 51

Table A.1. Calculated bending strength of concrete with a single rebar configuration (kip-inches)………… 63

Table C.1. Set 1 concrete cylinder compression strengths after 28 days (psi)……………………………….. 73

Table C.2. Set 2 concrete cylinder compression strengths after 28 days (psi)……………………………….. 73

Table C.3. Set 1 concrete cylinder compression strengths estimation after 14 days (psi)…………………… 74

Table C.4. Set 2 Concrete cylinder compression strengths estimation after 14 days (psi)…………………… 74Table C.5. Set one bending strengths using 14-day compression strengths (kip-inches)…………………….. 74

Table C.6. Bending strengths using 14-day compression strengths for single rebar (kip-inches)…………… 75

Table C.7. Bending strengths using 14-day compression strengths for double rebar (kip-inches)………….. 75

82

List of Equations Page #

Equation 5.1. .The buckling strength of steel plates…………………………………………………………. 14

Equation 5.2. .The expected bending strength of the steel plates…………………………………………….. 16

Equation 5.3. .Used to determine this relationship between M and _………………………………………… 42

Equation A.1. .The expected bending strength of the concrete and single rebar configuration……………… 63

Equation A.2. .The expected bending strength of the concrete and double rebar configuration……………… 64

Equation A.3. .The approximate shear strength of the concrete piers………………………………………… 65

Equation A.4.. The expected bending strength of wood……………………………………………………… 66

Equation A.5. The approximate shear strength of each wood column………………………………………. 67

Equation C.1. The compression strength of concrete at any given time…………………………………….. 73

Equation C.2. The 14 day compression strength of concrete………………………………………………… 74

Moment Resisting Wood Post-To-Concrete Peir Connection

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