The seismic performance of reinforced concrete

frames with wide band beams

J. S. Stehle, K. Abdouka, H. Goldsworthy, P. Mendis

The University of Melbourne, Australia, stehle@civag.unimelb.edu.au

Abstract

Experimental work on half scale interior and exterior subassemblages ispresented. The first series of tests are designed in accordance with Australiancode requirements for the non-ductile, ordinary moment resisting framecategory, and hence there are no special detailing requirements for seismicconsideration. The second series of tests are on joints with improved detailingwith the view to improve performance under cyclic loading. The results arecompared in the paper. The main difference between the detailing regimes is thatthe poorer detailing allows large crack widths to form due to torsion at the sideof the joint. The improved detailing prevents this from occurring. For theexterior connection it is achieved by concentrating the longitudinal beamreinforcement close to the column. For the interior connection a unique strategyof debonding longitudinal beam bars that pass well outside the joint core area isadopted. Preliminary time-history analysis results show that present constructiontechniques may be adequate for regions of low seismicity, however some cheapimprovements are suggested. For regions of higher seismicity, it is suggestedthat band beams may be used with the more ductile detailing as a primary orsecondary seismic resistance.

1 Design and details of the test specimens

The first series of tests included one interior and one exterior subassemblage of afour storey, six bay moment resisting frame (figure 1) detailed without anyspecial seismic considerations. Beam details of the first series of tests are given

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114 Earthquake Resistant Engineering Structures

in figures 2 and 3. All specimens were half scale. This was the largest possiblescale with the facilities and budget available.

Reinforcement detailing was determined purely from the strength demandsof the dead, live, wind and earthquake load combinations specified in theAustralian standards AS 1170.1,2 and 4, and the detailing requirements ofAS3600. It should be noted that consideration of earthquake loads had no impacton the design, other than for the provision of extra ligatures in the interiorcolumns for shear. It should be noted, however that a capacity design approachis not enforced by the codes of practice for this frame, hence a soft-storeycollapse mechanism is possible. Brittle shear failure in the columns may occurprior to this however because there is no strength hierarchy imposed for brittlefailures to be avoided. This is one example of a problem that may need to beaddressed in future code revisions.

The earthquake forces were determined using the simplified static method,with a peak ground acceleration of O.lg, a soil factor of 1.0 (rock), animportance factor of 1.0, a response modification factor of 4.0 (Rf=4.0 for non-ductile moment resisting frames) and a natural period determined from the codessimplified formula of height divided by 46 which for this frame is 0.31 seconds.The resulting force which is calculated can not be decreased by more than 20%using a dynamic analysis to calculate the period, even though the period of thestructure was found to be about 0.89 seconds using gross section stiffnesses.While this dynamic analysis did not account for the possibility of non-structuralcomponents adding stiffness, it was felt that using the full gross stiffnesses wasconservative enough. The effect of having an underestimated period led to theearthquake force being overestimated by about a factor of two. This is due to thelarge drop in the design acceleration response spectrum from a period of 0.31seconds to 0.89 seconds. This overstrength of two, coupled with otheroverstrengths such as that provided from a gravity load dominated design, thepresence of capacity reduction factors, and the fact that actual material strengthsare nearly always greater than the assumed design properties, all leads to a verylarge overstrength for the design level earthquake. In fact no damage may occurfor the design level earthquake and preliminary time-history analyses show thatthis is true. This bears well for current design practice, however, due to thelimited knowledge of seismicity in Australia, what constitutes the design levelearthquake is subjective, hence it is felt that it would be prudent to provide atleast some very basic and cheap improvements to the current detailingrequirements. An example of this would be to provide more ligatures in columnsin case a soft-storey were to occur.

For the second series of tests, which again included one interior and oneexterior subassemblage, it was decided that a more ductile response should beaimed for with minimal detailing change. The main changes (see figures 2 and 3)that were made were to avoid torsional cracks and bottom bar pullout, significantproblems observed in the first series of tests. Bottom bar pullout is easilyprevented. For the exterior specimen, the bottom bars are cogged upwards at the

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Earthquake Resistant Engineering Structures 115

back face of the spandrel beam. For the interior specimen, the bottom bars aresimply made continuous.

For the exterior specimen, torsional cracking can be prevented byconcentrating a sufficient amount of beam longitudinal bars near the column, sothat the torsional cracking strength of the spandrel beam cannot be reached.

For the interior specimen, concentrating the longitudinal beam bars was notconsidered to be a good idea for two reasons - there is too much congestion closeto the column, and an undesirable strong beam - weak column strength hierarchyis more likely. The idea that the first author thought of was to debond the beamlongitudinal bars that pass far outside the joint core. The debonding processallows the bars not to develop any out of balance moment on the joint, since abar can not be in compression at one end and in tension at the other end withouthaving bond resistance to transfer the forces from the steel to the concrete. Asinsufficient out of balance moment can be developed to reach the torsionalcracking moment, torsional cracks will not occur. In effect, the effective widthof the beams for strength for out of balance moments is reduced, ensuring adesirable strong column - weak beam strength hierarchy. Consideration of thissystem under balanced moments, such as that existing for gravity loading, is alsorequired. The debonded bars simply have equilibrated tension in the top bars andcompression in the bottom bars. For the transfer of shear, the effective width ofthe beam is reduced though, so the reduced width of the beam requires someligatures. As the effective shear width will gradually spread out to the full widthof the beam for sections further towards mid-span, it seems prudent to graduallyto increase the ligature spacing, and have the ligatures extend across the beam'sfull width. For the sake of uniformity ligatures across the whole width of thebeam were used, even though they are not really required in the outer portions ofthe beam beside the joint core, in the debonded bar zone. It was by coincidencethat the ligature spacing matched that proposed by Quintero-Febres and Wight[2] for wide beams detailed for high seismicity. The debonding idea andextensions of it to other interior connections, such as narrow concrete beam,steel and composite joints, is the subject of a current patent application.

Exteriorsubassemblage

/Interior subassemblage

\c xs c —t —

//

-<— 2

: 3.4m; 3.4m; 3.4m; 4.2m

8.4m 9.6m 9.6m 9.6m 9.6m 8.4m

Bay width = 8.4 m in out of plane direction.Beams: 2400mm wide, 400mm deep

Columns: 500mm squareEdge beams: 300mm wide, 600mm deep

Figure 1: Full scale geometry

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116 Earthquake Resistant Engineering Structures

L-8-Y12PLAN ( Top Layer)(Slab reo RIO @ 75 top - not shown)

tip

L-8-Y12 I-8-Y12

PLAN ( BOTTOM LAYER)

HJTo

PLAN < Top Layer)(Slab reo RIO @ 75 top - not

llllllTT

PLAN ( BOTTOM LAYER)

Figure 2: Details of half scale interior specimens (left - INT-1, right - INT-2)

PLAN C Top Reinforcement)

Figure 3: Details of half scale exterior connections (left - EXT-1, right - EXT-2)

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Earthquake Resistant Engineering Structures 117

2 Experimental setup

The specimens were tested in a reaction frame built specifically for this program.The frame was braced against lateral sway and firmly secured to the stronglaboratory floor.

The ends of the beam and the columns of the specimens were pin-endedrepresenting the approximate points of contraflexure for lateral loading in thereal structure. The effects of the gravity loads were modelled using acombination of concrete blocks on the beam and axial compressive forces on thecolumns. Jacking prestressing strands against the ends of the columns providedthe axial column load. Axial forces of 300 kN and 600 kN for the exterior andinterior specimens respectively on the column were one-fourth the axial columnload of the frame due to Dead +0.4 Live loads. Concrete blocks hung from thebeam were used to model the bending moment and the shear force in themembers prior to the application of the lateral loads. A hydraulic actuatorconnected to the pin end of the lower column provided the lateral loadsimulation that was in the form of displacement controlled quasi-static loading ofincreasing drift ratios. Drift ratio is defined as the ratio of the relativedisplacement of the bottom to top pins divided by the distance between these twopins. The test specimens were extensively instrumented with strain gauges,displacement transducers and load cells. The data acquisition system included anOrion Data Logger, a Houston Instruments X-Y plotter and a 486-IBMcompatible PC.

3 Test results

The hysteretic responses of the test assemblages are presented in figures 4.1, 4.2and 5.1, 5.2 for interior and exterior subassemblages respectively. In general thefirst series of tests, detailed as non-ductile moment resisting connections,exhibited a number of brittle failure mechanisms. The second series of tests, withimproved detailing, avoided brittle failure mechanisms to a large extent. In alltests, it should be noted that a 20mm displacement was achieved in the elasticrange of response. This converts to a 40mm drift at full scale. Given thatearthquakes with a peak ground displacement in the order of 20mm for rock sitesare expected in Australia for the one in 475 year design event [4], it appears thatthis form of construction has no problems in Australia, except perhaps for shearin the columns. Further time-history analysis, accounting for soil effects isrequired to confirm this. The performance of this form of construction in regionsof higher seismicity is currently being evaluated in ongoing research at theUniversity of Melbourne.

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118 Earthquake Resistant Engineering Structures

3.1 Specimen INT-l

The first interior connection achieved a hysteretic response (see figure 4.1)which looks practically elasto-plastic though with significant stiffnessdegradation. This appears to be a satisfactory response, however, considerationof the type of damage which occurred is required.

Firstly, damage occurred mainly in the beam, with the bottom columndisplaying only minor flexural cracking. This is despite the beams having greaterflexural capacity, based on their full width, than the columns. The column tobeam moment capacity ratio was calculated as 0.98 using measured materialproperties, rectangular stress block theory, by assuming that no bar slip ordegradation would occur in the joint core, using 0.4 times the bar yield stress incalculating beam positive moments to account for poor bar anchorage, and byassuming the full beam width is effective. To account for the reduction of beamwidth due to torsional cracking, the column to beam moment capacity ratio wascalculated based on the width of beam framing directly into the column and thetorsional cracking strength of the beam on the sides of the column using Hsu'sformula [3] which is recommended by Gentry and Wight [1]. Using this methoda value of 1.44 was obtained.

On inspection of the specimen during testing, torsional cracks wereobserved to form at a drift level of 30mm (1.6% drift). These cracks becamevery wide (2mm approx.) in later cycles. This behaviour was deemedunsatisfactory, since the punching shear strength for the ultimate gravity loadcombination is compromised. It should be noted that no drop in strengthoccurred after the formation of torsional cracks during the test. It is believed thatthere is sufficient aggregate interlock along the crack for some torsional strengthto be maintained. This belief is supported by observation of strain gaugereadings in the outer beam bars. They show that there is still some out-of-balance moment in the beam outside the torsional crack region. Some finiteelement work has also been conducted using ANSYS version 5.4 and itsSOLID65 reinforced concrete element. Although, not presented here, it has beenfound that the factor specified for the shear transfer across open cracks toaccount for aggregate interlock is an important factor.

Other brittle failure mechanisms did not really diminish the performance ofthe connection to a degree that would cause concern. It was noted however thatthe bottom beam bars did not reach their full yield capacity. The strain on 10bottom bars were recorded at distances of 120mm and/or 320mm from the endof the bar (The end of the bar terminating 120mm past the column face). Twelveof the fourteen strain gauges in these locations read maximum tensile strains of1200 microstrain or less (yield at 2100 microstrain approx.). The other twogauges had maximums of 1500 and 1900 microstrain. For improvedperformance it was felt that it would bear little cost to increase the anchorageprovided, hence this modification was made for the second test. Strain gaugereadings of the longitudinal column bars passing through the joint core also

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Earthquake Resistant Engineering Structures 119

showed some bar slip taking place at higher drift levels. This effect was nottreated in the second test, as the first test showed little column strengthdegradation.

3.2 Specimen INT-2

The hysteretic response of the second interior specimen (figure 4.2) is fairlysimilar to that achieved for the first test, except a lower yield strength and ahigher maximum strength are attained. The lower yield strength can be attributedto the debonding of the outer beam bars since the torsional cracking capacityisn't developed. The higher maximum strength can be attributed to the fullyanchored bottom bars.

The primary damage that occurred was beam flexural cracking, however thecrack widths were fairly small - about 0.3mm at the maximum 66.5mm drift(3.5% drift), which is not bad in terms of serviceability. No torsional cracksformed. The column to beam moment strength ratio was controlled to 1.96. Thisensured the strong column - weak beam strength hierarchy.

From this test, it was concluded that the bar debonding idea was a completesuccess.

3.3 Specimen EXT-1

This was the first specimen to be tested and as mentioned earlier no a-seismicdetailing was provided. Beam longitudinal reinforcement was placed uniformlyacross the width of the beam. Bottom beam bars ended with a straight baranchorage length of a nominal 10 bar diameter into the support.

Referring to figure 5.1 the specimen showed two different types ofbehaviour in the two different directions of loading. There was a significant andsharp reduction in the lateral load resistance in the positive direction (top ofbeam in compression) after the maximum resistance was reached in this loadingdirection. The longitudinal bottom beam bars pulled through the concrete in thespandrel beam and large cracks appeared in the interface between the wide beamand the spandrel beam, thus leading to the reduced lateral resistance.

In the negative direction of loading (top of beam in tension) the behaviourof the beam was superior to its behaviour in the positive loading direction.Lateral load resistance was higher and the reduction of the lateral resistance withincreased drift ratios was more gradual. However the maximum lateral loadobtained from the test was less than the "upper bound" lateral load capacitybased on the full contribution of all the top bars in the wide beam. This was dueto the fact that once the torsional element cracked (which occurred at a drift ratioof approximately 1.6%) the top bars anchored in it lost their anchorage and werenot capable of transferring additional moments to the column through torsion inthe spandrel beam. The cracking in the spandrel beam extended and widened asthe drift level increased during the test and additional cracks associated with

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120 Earthquake Resistant Engineering Structures

positive direction loading developed. There was one major crack in the beam atthe face of the column at which the inelastic behaviour was concentrated with avery large crack width. Strain gauges measuring the strain on the longitudinalreinforcement at the location of the crack indicated very high strain levels.

At large levels of drift (3.5%) the crack in the spandrel beam became solarge that the specimen behaved basically as two independent parts. It is believedthat a limited level of aggregate interlock and the contribution of the spandrelbeam reinforcement is the reason behind the residual strength of the specimen atthis high level of drift.

3.4 Specimen EXT-2

The overall load versus displacement curve is shown in figure 5.2 The specimenbehaved much better than EXT-1 in the way that the strength in both directionsdid not deteriorate. The theoretical strength of the beam in the negative directionwas exceeded which can be attributed to the strain hardening of a number of topbeam bars. The torsional crack at the spandrel beam started tight at a higher driftlevel than the first torsional crack of EXT-1 and maintained its tightnessthroughout most of the test. The bottom longitudinal bars did not pull outbecause of their improved anchorage. A number of flexural cracks developed onthe top and bottom surfaces of the beam and thus resulted in smaller crackwidths unlike the undesirable major crack that opened wide in EXT-1.

o<

-100Displacement (mm)

Figure 4.1: INT-1 hysteresis

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Earthquake Resistant Engineering Structures 121

o<

T-50 -

-75

-100

Displacement (mm)

Figure 4.2: INT-2 hysteresis

20 40 60 80

T3COO^ -100 -80o(0

60 80

Displacement (mm)

Figure 5.1: EXT-1 hysteresis

40

50 100

-80 -LDisplacement (mm)

Figure 5.2: EXT-2 hysteresis

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122 Earthquake Resistant Engineering Structures

4 Conclusions

(1) Band beam construction in Australia appears to be able to withstand code-specified design level earthquakes with little, if any damage, although furtheranalysis for the performance on soil sites is required.

(2) The following minor detailing changes are recommended for future bandbeam ordinary moment resisting frames:

- ligatures in columns to be provided to withstand the shear that may developin a column in a soft-storey collapse.

• bottom longitudinal beam bars to be anchored sufficiently to have theiryield stress developed at the centreline of a support.

- beam longitudinal bars in an exterior connection to be placed across thewidth of the band beam such that the moment transferred by the bars anchoredoutside the column core does not exceed the torsional strength of the spandrelbeam.These changes are intended to improve the frame's ductility at a very minorextra expense, allowing improved performance in earthquakes which may bemore extreme than the code design level.(3) Band beam construction may be used successfully in regions of higherseismicity, in particular, with the use of the bar debonding idea (subject to patentapplication) tested in the second interior specimen. Significant displacementcapacity is obtainable, hence where drift or P-delta effects are of concern, it maybe prudent to use a band beam moment resisting frame as a secondary seismicresistance with shear walls or a perimeter frame as the primary lateral loadresistance. Further research is required for regions of higher seismicity.

5 References

1. Gentry, T.R. and Wight, J.K., Reinforced Concrete Wide Beam-ColumnConnections Under Earthquake-Type Loading. Report No. UMCEE92-12.The University of Michigan Ann Arbor USA, 1992.

2. Quintero-Febres, C. G. and Wight, J.K. Investigation on The SeismicBehaviour ofRC Interior Wide Beam-Column Connections. Report No.UMCEE 97-15. The University of Michigan Ann Arbor USA, 1997.

3. Hsu, T.T., Torsion of Structural Concrete-Plain Concrete RectangularSections. Torsion of Structural Concrete, American Concrete Institute SP-30. Detroit Michigan, pp. 203-238, 1968.

4. Edwards, M., Wilson, J. ,Lam, N. and Hutchinson, G. The DisplacementBased Approach From An Intraplate Perspective. Australasian StructuralEngineering Conference Auckland, pp. 713-720, 1998.

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