Geotechnical Hazards Assessment Report

Prepared for

Poseidon Resources5780 Fleet Street, Suite 140

Carlsbad, CA 92008

Geotechnical Hazards Assessment Report

Huntington Beach Seawater Desalination Project

Huntington Beach, California

Prepared by

595 Market Street, Suite 610 San Francisco, CA 94105

Project Number WG1665

March 2013

Geotechnical Hazards Assessment Report

Huntington Beach Seawater Desalination Project

Huntington Beach, California

This report was prepared under the supervision and direction of the undersigned.

Prepared by:

Geosyntec Consultants, Inc.

595 Market Street, Suite 610 San Francisco, CA 94105

Neven Matasovic, Ph.D., P.E., G.E. Jennifer Donahue, Ph.D., P.E. Associate Project Manager

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TABLE OF CONTENTS

EXECUTIVE SUMMARY .............................................................................................. 1

1. INTRODUCTION ............................................................................................. 1-1

2. BACKGROUND AND SCOPE-OF-WORK .................................................... 2-1

2.1 Project Background .................................................................................. 2-1 2.2 Geosyntec Scope-of-Work ....................................................................... 2-2

3. FIELD INVESTIGATION ................................................................................ 3-1

3.1 Field Investigation .................................................................................... 3-1 3.2 Review of Previous Field Investigations .................................................. 3-1

4. SUBSURFACE CONDITIONS ........................................................................ 4-1

4.1 General Geologic Conditions ................................................................... 4-1 4.2 Subsurface Soil Conditions ...................................................................... 4-1 4.3 Groundwater Conditions ........................................................................... 4-2

5. DESIGN BASIS ................................................................................................ 5-1

6. SEISMIC HAZARD ANALYSIS ..................................................................... 6-1

6.1 Seismic Hazard Analysis in support of Design ........................................ 6-1 6.2 Discussion on the Magnitude assigned to the Newport-Inglewood Fault 6-2 6.3 Supplemental Seismic Hazard Analysis – Parametric Study ................... 6-2

6.3.1 Moment Magnitude for the Newport-Inglewood Fault ................ 6-3

6.3.2 Site-to-Source Distance ................................................................ 6-3

6.3.3 Parametric Study Results ............................................................. 6-3

7. EVALUATION OF DESIGN GROUND MOTIONS ...................................... 7-5

8. SITE RESPONSE ANALYSIS ......................................................................... 8-1

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9. SURFACE FAULT RUPTURE HAZARD ....................................................... 9-1

9.1 Likelihood of Surface Fault Rupture at the Site ....................................... 9-1 9.2 Surface Fault Rupture Finite Element (FE) Analysis ............................... 9-2

9.2.1 Model Geometry .......................................................................... 9-2

9.2.2 Material Properties ....................................................................... 9-3

9.2.3 Magnitude of Model Fault Displacement .................................... 9-3

9.2.4 Structural Damage Thresholds ..................................................... 9-4

9.2.5 Model Results .............................................................................. 9-4

9.3 Fault Rupture Conclusion ......................................................................... 9-5

10. LIQUEFACTION AND LATERAL SPREAD HAZARD ASSESSMENT ... 10-6

10.1 Liquefaction Triggering .......................................................................... 10-6 10.2 Effects of Liquefaction ........................................................................... 10-7

10.2.1 Residual Shear Strength of Liquefied Soil ................................. 10-7

10.2.2 Liquefaction Induced Settlement ............................................... 10-7

10.2.3 Evaluation of Lateral Spreading Displacements ........................ 10-8

10.3 Liquefaction and Lateral Spread Mitigation Strategies ........................ 10-12

11. TSUNAMI HAZARD ASSESSMENT ........................................................... 11-1

11.1 Baseline Tsunami Water Level ............................................................... 11-1 11.2 Projected Sea Level Rise over Design Life ............................................ 11-2 11.3 Sea Level Rise-Adjusted Tsunami Water Level ..................................... 11-3 11.4 Tsunami Inundation Potential ................................................................. 11-3 11.5 Tsunami Impact at the Site ..................................................................... 11-3 11.6 Tsunami Impact Mitigation Strategies ................................................... 11-4

12. RECOMMENDATIONS ................................................................................. 12-1

13. LIMITATIONS ................................................................................................ 13-1

14. REFERENCES ................................................................................................ 14-2

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LIST OF TABLES

Table 10-1: Liquefaction Induced Settlement

Table 10-2: LDI-Based Lateral Spread Displacement

Table 10-3: Newmark-Based Lateral Spread Displacement

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LIST OF FIGURES

Figure 2-1: Site Layout

Figure 4-1: Geologic Map

Figure 4-2: Geologic Cross Section

Figure 4-3: Idealized Soil Profile

Figure 6-1: Deaggregated Seismic Hazard for the Site

Figure 6-2: Acceleration Response Spectrum – Bedrock

Figure 6-3: Results of Deterministic Seismic Hazard Analysis for Mw 7.5

Figure 7-1: Design Ground Motions

Figure 8-1: Modulus Reduction and Damping Curves

Figure 8-2: Acceleration Response Spectra – Surface Response

Figure 8-3: Calculated Peak Shear Stress Profile

Figure 9-1: Cross Section A-A'

Figure 9-2: Fault Rupture Impact Simulation - Finite Element Model

Figure 9-3: Main Fault Trace Displacement Magnitude

Figure 9-4: Secondary Fault Displacement Magnitude

Figure 9-5: Fault Rupture Impact Simulation - Results

Figure 10-1: Site Layout with Stability Cross-Sections

Figure 10-2: Evaluation of Liquefaction Triggering

Figure 10-3: Post-Liquefaction Idealized Soil Profile

Figure 10-4: Evaluation of Post-Liquefaction Strength

Figure 10-5: Liquefaction Induced Settlement

Figure 10-6: LDI-Based Lateral Spread Displacement

Figure 10-7: Section SA-SA’ - Shallow Pseudo-Static Stability Surface

Figure 10-8: Section SB-SB’ – Shallow Pseudo-Static Stability Surface

Figure 10-9: Section SC-SC’ – Shallow Pseudo-Static Stability Surface

Figure 10-10: Section SB-SB’ – Deep Pseudo-Static Stability Surface

Figure 11-1: Tsunami Hazard Analysis Key Terms

Figure 11-2: Baseline Tsunami Water Level Elevation

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LIST OF FIGURES (CONT’D)

Figure 11-3: Projected Sea Level Rise over Design Life

Figure 11-4: Proposed Finished Floor Elevations and Locations of Existing Berms

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LIST OF APPENDICES

Appendix A: Site Investigation Data

A-1: Geosyntec Cone Penetration Testing Data

A-2: GLA [2002] Investigation Data

A-3: Ninyo & Moore [2011] Investigation Data

Appendix B: Site Response Analysis

Appendix C: Slope Stability Analysis Results

Appendix D: Evaluation of CCC Staff [July 13, 2012] Tsunami Recommendations

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EXECUTIVE SUMMARY

The Huntington Beach Seawater Desalination Project (Project) calls for construction of a desalination plant with an output capacity of approximately 50 million gallons per day. The plant is planned to be located within an approximately 12-acre site (Site) at 21730 Newland Street in the City of Huntington Beach, California, adjacent to the AES Corporation (AES) Huntington Beach Generating Station.

In order for the Project to proceed, a Coastal Development Permit must be obtained from the California Coastal Commission (CCC). A permit application was submitted, and CCC [July 13 2012] responded with a Notice of Incomplete Coastal Development Permit Application. This notice requested additional information in two main categories: (i) Geologic Hazards and Project Stability; and (ii) Tsunami Hazards and Risks. This Geotechnical Hazards Assessment Report is in response to the CCC’s response letter [CCC, 2012], dated 13 July 2012, which comments on the Coastal Development Permit for the subject site, as well as subsequent requests by the Commission staff for additional hazard analysis1.

The findings presented here supersede those presented in Geosyntec’s previous Geotechnical Hazards Assessment Report dated December 2012 [Geosyntec, 2012]

Geologic Hazards and Stability

Surface Fault Rupture and Structural Stability

The SEIR and other submitted documents identified a potential for subsurface faulting beneath the proposed project site. Mitigation Measure GEO-l of the SEIR required Poseidon to conduct a fault hazard investigation to evaluate the location and extent of faulting (if any) and to identify the potential for fault rupture propagation to ground surface.

1 The Report also includes additional analysis as requested by Commission staff in its February 19, 2013 and March 1, 2013 emails and the February 27, 2013 meeting among Poseidon, Geosyntec and Commission staff.


Finding

Geosyntec performed a review of the available literature to assess the likelihood of a subsurface fault rupture at the Site and found such fault rupture to be unlikely. In addition, Geosyntec performed a field investigation consisting of 5 Cone Penetration Test soundings, ranging from depths of 50 ft to 98 ft below ground surface. This information, coupled with the previous investigations and literature review was used to develop a Finite Element model to evaluate the impacts of a fault rupture below the Site, in this case to represent the postulated location of the South Branch Fault. The structure was conservatively assumed to be built on a concrete mat foundation. A soil zone was modeled between the mat foundation and the estimated top of bedrock. The elevation of the top of bedrock was estimated at -200 ft above mean sea level (ft msl) based on a geologic cross section. The evaluation was performed using conservative assumptions (e.g., the postulated fault was assumed to be located directly below the Site), in order to assess a “worst-case” scenario (i.e., conditions as severe or more severe than considered likely at the Site). The estimated displacement on the postulated fault at the top of the bedrock was approximately 0.95 feet and would occur during a magnitude Mw=7.1 event on the main trace of the Newport-Inglewood fault system.

The results of Geosyntec’s Finite Element model evaluation indicate that the approximately 200-ft thick deposit of alluvial sediments below the Site mitigates the fault rupture hazard. The results further indicate that the maximum model-calculated angular distortion is approximately 1/276. This model-calculated value is more severe than the Serviceability Limit State (associated with architectural damage, such as wall cracking) of 1/300 but is less severe than the Ultimate Limit States (associated with structural damage, such as frame cracking) of 1/170. This result suggests that if the postulated fault rupture were to occur beneath the project site, the proposed structures may experience repairable aesthetic and temporary serviceability issues, but significant structural damage is unlikely. Based on results of the investigation and analysis, no changes to the project layout, design, engineering or mitigation measures are needed to improve structural stability against fault rupture.

Lateral Soil Spread

The SEIR identified the potential for lateral soil spread at the project site and required through Mitigation Measure GEO-2 that Poseidon conduct a Site-specific geotechnical investigation. Further, Poseidon was to provide the results of the geotechnical


investigation and subsequent study. Lastly, the CCC [July 13, 2012] letter requested any proposed changes to the project layout, along with associated changes to the design, engineering, mitigation, and other measures needed to avoid the risk of lateral soil spread and to ensure structural stability.

Finding

Geosyntec performed a state-of-the-practice lateral spread analysis using two methods: (i) the strain potential approach after Zhang et al. [2004]; and (ii) the Newmark sliding block approach after Bray and Travasarou [2007]. The Huntington Beach Flood Control Channel, which serves as a free face for lateral spread development, is located approximately 150 feet from the nearest proposed on-Site structure. For the purposes of this analysis, the sheet pile wall acting as the flood protection sidewall for the Huntington Beach Channel was conservatively assumed to provide no structural support against lateral spreading during a seismic event. For a distance of 150 feet from the free face, the lateral spread displacement on the project site is estimated to range from approximately 15 to 38 inches. This range of displacement can be accommodated through design features as described below.

Based on results of the investigation and analysis, no change to the project layout is needed for structural stability. However, to accommodate the potential lateral spread displacement on the Site, Geosyntec recommends that, during the Project’s design phase, the Project’s Structural and Geotechnical Engineers collaborate on the design of a foundation system for the proposed structures that can accommodate lateral spread displacement of approximately 15 to 38 inches. Such foundation system may include both geotechnical ground improvement methods to reduce displacements to acceptable levels, and structural design methods to allow Site structures to tolerate the estimated displacements. This recommendation is incorporated in Design Measure B, presented under the “Liquefaction” discussion below.

"Design-Level" Earthquake

The CCC [2012] letter requested additional information on the "design-level" earthquake based on Site-specific geotechnical/geophysical investigations and consistent with the most recent update of the Uniform California Earthquake Rupture Forecast. Poseidon was also asked to provide seismic design parameters corresponding to the most recent update of the California Building Code and describe the maximum


credible earthquake and peak ground acceleration at the proposed project site. This information will be used to provide the structural design parameters Poseidon will use to design the proposed on-Site structures to withstand the “design-level" seismic forces.

Finding

Geosyntec performed a site-specific probabilistic seismic hazard analysis, developed design ground motions, and performed site response analysis. The project should be built in accordance with the CBC [2010] requirements; hence, the “design-level” earthquake is an event with 2 percent probability of exceedance in 50 years (2% PE in 50 years). With a Moment Magnitude (Mw) of 7.1 assigned by the United States Geological Survey and a site-to-source distance of 0.5 miles, the Newport Inglewood Fault is the governing seismic source for this site. Deterministic sensitivity analyses were performed on the magnitude and distance parameters of the fault and found that the probabilistic seismic hazard analyses are reasonably conservative. The corresponding bedrock Peak Horizontal Ground Acceleration (PHGA) for this fault is of 0.61 g.

Because the Site is considered to be vulnerable to liquefaction, the Site is considered to a Site Class F per the CBC [2010], and therefore requires a site-specific seismic response analysis. One-dimensional non-linear seismic site response analyses of the Site were conducted using the computer program D-MOD2000 with five representative accelerograms scaled to the design bedrock PHGA of 0.61 g. Per CBC [2010] requirements, the average acceleration response spectrum calculated from the site-specific response analysis should be compared against a code-based minimum. This code-based minimum spectrum corresponds to 80% of the probabilistically-established acceleration response spectrum for a Site Class E site.

The average acceleration response spectrum from the site-specific response analysis was less than the code-based minimum. As such, the following design measure is recommended:

• Design Measure A: Geosyntec recommends the Project be designed following CBC [2010] requirements for Site Class F, with an acceleration response spectrum corresponding to 80% of the Site Class E response spectrum.


Liquefaction

Poseidon was requested to identify the areas and depths of soils on Site with liquefaction potential that could affect the location or design of project components and identify proposed changes to the project layout, along with associated changes to the design, engineering, mitigation, and other measures needed to mitigate the risk of liquefaction and associated effects on the proposed structures.

Finding

A supplemental field investigation was performed to provide enhanced information with regard to the subsurface conditions. Five Cone Penetration Test (CPT) soundings were performed at representative locations of the Site. Two liquefiable zones were identified in the subsurface soils: (i) an upper layer approximately 4 feet-thick; and (ii) multiple lenses between 45 and 70 feet below ground surface (bgs). The presence of these potentially liquefiable soil lenses below 45 ft bgs were also observed in Geosyntec’s review of CPTs from previous investigations by others on and near the Site. Based on Geosyntec’s evaluations, it is estimated that up to 9 in. of total liquefaction-induced reconsolidation settlement may occur at the Site. This level of liquefaction-induced settlement is within the normal range for a Site with this type of soil profile in an area of high seismicity.

Geosyntec recommends that, during the Project’s design phase, the Project’s Structural and Geotechnical Engineers collaborate on the design of a foundation system for the proposed structures that can accommodate approximately 9 inches of liquefaction-induced settlement. Such foundation system may include both geotechnical ground improvement methods to reduce settlements to acceptable levels, and structural design methods to allow Site structures to tolerate the estimated settlements. This recommendation along with Geosyntec’s recommendation to accommodate potential lateral soil spread on the Site has been incorporated into recommended Design Measure B:

• Design Measure B: During the Project’s design phase, Geosyntec recommends that the Project’s Structural and Geotechnical Engineers collaborate on the design of a foundation system for the proposed structures that can accommodate (1) approximately 9 inches of liquefaction-induced settlement; and (2) lateral spread displacement of approximately 15 to 38 inches. Such foundation system


may include both geotechnical ground improvement methods to reduce the anticipated settlements and displacements to acceptable levels, and structural design methods to allow Site structures to tolerate the estimated settlements and displacements.

Tsunami Hazards and Risks

The CCC [2012] letter states that the 2010 SEIR used studies published in 1985 and 1996 to conclude that maximum tsunami heights of 7.5 feet above mean sea level would result in the proposed project being at low risk of tsunami-related hazards. However, the CCC [2012] letter states that more recent studies suggest the proposed project would be at much higher risk, with expected maximum tsunami heights of 16.0 feet and a tsunami run-up zone extending more than a mile inland of Poseidon's proposed site. The CCC [2012] letter also states that the 2012 National Academy report identifies an anticipated sea level rise of up to about three feet over the expected life of the proposed development. Therefore, the CCC [2012] letter suggests that Poseidon's updated tsunami hazard assessment should be based on an expected maximum 16-foot tsunami run-up height and should incorporate the additional height needed to reflect high tide levels and a three-foot sea-level rise.

Finding

The conclusions regarding tsunami hazard risk found in these reports are generally consistent with the conclusions reached by the City of Huntington Beach in the project’s Final SEIR – i.e., that tsunami hazard risk would be low. As described in detail in Appendix D of the Report, Geosyntec does not consider these CCC [July 13, 2012] recommendations to be adequately supported. Therefore, Geosyntec considers additional evaluation to be needed.

According to the NRC [2012] report, the “upper-bound” sea level rise projection in the Los Angeles, California, area is approximately 2.0 ft in the year 2050 (i.e., the final year of the planned design life of the proposed facility). Based on this projection, Geosyntec considers 2.0 ft to be a conservative value of sea level rise for initial analysis, design, and construction.

Based on the location of the mapped tsunami inundation line identified by Cal EMA [2009], a baseline tsunami water level elevation of 10 ft msl is considered appropriate


for the Site. As indicated in the Cal EMA [2009] map, this value conservatively accounts for “mean high water” sea-level conditions. Adding the “upper-bound” projected sea level rise of 2.0 ft to the baseline tsunami water level elevation of approximately 10 ft msl gives a sea level rise-adjusted tsunami water level elevation of approximately 12 ft msl.

Site-Specific Information

The CCC [2012] letter requested that Poseidon provide surveyed elevations (above mean sea level) of existing site features, of proposed changes to those features, and of the proposed facility components. This should include elevations of existing grades within the proposed facility footprint, including existing on-site berms, and proposed final elevations of all project components, including buildings, tanks, pumps, and any proposed changes to berms, flood control features of the adjacent flood channel, and other on- and off-site project elements. The letter also asked Poseidon to identify the amounts and locations of fill needed to attain the proposed elevations.

Finding

Finished floor elevations for proposed Site improvements are shown in Figure 11-4 of this report. The proposed finished floor elevations range from approximately 9.0 ft msl to approximately 14.0 ft msl.

Use of Updated Studies and Information:

The CCC [2012] letter requested that Poseidon assess the tsunami hazards based on more recent studies than those used in the project SEIR and the City's CDP.

Finding

Geosyntec [2013] reviewed the five references cited by CCC [July 13, 2012].

Four of these references (i.e., County of Orange [2006, 2008], Cal EMA [2009], and City of Huntington Beach [2011]) pertain to possible tsunami hazard. However, based on Geosyntec’s [2013] review, these references were found to be inadequate for evaluation of tsunami hazard at the Site.


The remaining reference (i.e., NRC [2012]) pertains to sea level rise over the design life of the proposed facilities. Based on Geosyntec’s review, this report suggests an “upper-bound” sea level rise of approximately 2.0 ft over the planned design life of the proposed facility.

Geosyntec performed additional review of the relevant technical literature and performed evaluations to assess the potential tsunami impacts at the Site. Based on these evaluations and considering mean high water level, sea level rise over the design life, and a tsunami inundation area as identified by Cal EMA [2009] Geosyntec estimated tsunami water level elevation as approximately 12 ft above mean sea level (ft msl). Based on proposed finished floor elevations ranging from approximately 9.0 ft msl to approximately 14.0 ft msl, tsunami inundation depth is estimated to be up to approximately 3 ft for the tsunami event considered.

Design Modifications and Mitigation Measures

Based on the updated site-specific information and tsunami assessment, the CCC [2012] letter requested that Poseidon identify the design modifications and mitigation measures that should be employed to avoid and reduce tsunami-related risks. The letter stated that this should include the expected water elevations serving as the basis for the assessment and should describe the method, location, and design basis for any structural components - e.g., the engineering strength of proposed structural components, consistency with the UBC requirements.

Finding

Based on the results of Geosyntec’s evaluations, it appears likely that portions of the Site could be inundated during the tsunami event considered. Geosyntec considered potential tsunami impacts and recommended mitigation measures to reduce tsunami-related hazards.

Possible tsunami impacts at the Site related to inundation of up to approximately 3 ft of water include seepage, soil erosion, and loading on proposed structures. The impact of seepage is anticipated to be small, as water inundation will be temporary. The soil erosion impact also is likely to be small as much of the Site is anticipated to be covered with concrete or asphalt pavement. Tsunami-related loading on the proposed structures can be mitigated by structural design. Geosyntec has recommended the following


design measures to accommodate potential risks from tsunamis and tsunami-related loading:

• Design Measure C: Geosyntec recommends the implementation of SEIR mitigation measure HWQ-3: Prior to issuance of grading permits, the applicant shall submit to the City for approval a plan outlining the specific planning measures to be taken to minimize or reduce risks to property and human safety from tsunami during operation. Planning measures could include but would not be limited to the following: (a) Provision of tsunami safety information to all facility personnel, in addition to posting signage on site; (b) identification of the method for transmission of tsunami watch and warnings to facility personnel and persons on the site in the event a watch or warning is issued; and (c) identification of an evacuation site for persons on site in the event of a tsunami warning;

• Design Measure D: Geosyntec recommends the development of a coordinated Emergency Response Plan with AES HBGS prior to the commencement of Project operations; and

• Design Measure E: Geosyntec recommends the incorporation of tsunami-resistant design features into the design of proposed structures that are sufficient to accommodate potential inundation of up to approximately 3 ft. of water. Guidance on tsunami-resistant design that can sufficiently accommodate these inundation levels and provide for vertical evacuation, if necessary, is available in the Applied Technology Council report titled Guidelines for Design of Structures for Vertical Evacuation from Tsunamis [ATC, 2008].

Site tsunami hazard is not anticipated to present a significant risk to public health and safety. According to the project’s SEIR, there will be fewer than twenty personnel on site during desalination plant operations. It is noted that the Orange County Grand Jury report on Tsunami Hazards [2008]2 finds that the City of Huntington Beach has one of the most advanced tsunami early-warning systems in the County. The City of

2 Orange County Grand Jury [2008] “PARADISE LOST: If a Tsunami Strikes the Orange Country Riviera…”


Huntington Beach is also developing a Local Hazard Mitigation Plan3 that addresses tsunami risk to public health and safety.

3 City of Huntington Beach [2011] “Draft, Hazard Mitigation Plan, City of Huntington Beach”

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1. INTRODUCTION

Geosyntec Consultants, Inc. (Geosyntec) is pleased to provide Poseidon Resources (Poseidon) with this Geotechnical Hazards Assessment Report. This Report summarizes the findings of Geosyntec’s field investigation and geotechnical evaluations (i.e., seismic hazard assessment, soil liquefaction potential evaluation and assessment of its impacts fault rupture propagation potential evaluation), and evaluation of site-specific tsunami hazard. These evaluations were performed at Poseidon’s request in response to the California Coastal Commission staff’s (CCC Staff) response letter [CCC Staff, 2012], dated 13 July 2012, which comments on the Coastal Development Permit for the subject site. The Report also includes additional analysis as requested by CCC Staff in its February 19, 2013 and March 1, 2013 emails to Poseidon Resources and from the February 27, 2013 meeting involving Poseidon, Geosyntec and CCC Staff.

This Report was prepared by Jennifer Donahue, Ph.D., P.E., Brian Martinez, Ph.D., and Alan F. Witthoeft, P.E. of Geosyntec. Christopher Hunt, Ph.D., P.E., G.E. and Neven Matasovic, Ph.D., P.E., G.E., also of Geosyntec, reviewed this Report in accordance with Geosyntec’s internal peer review policy.

This Report supersedes Geosyntec’s previous Geotechnical Hazards Assessment Report dated December 2012 [Geosyntec, 2012].


2. BACKGROUND AND SCOPE-OF-WORK

2.1 Project Background

The Huntington Beach Seawater Desalination Project (Project) calls for construction of a desalination plant with an output capacity of approximately 50 million gallons per day. The plant is planned to be located within an approximately 12-acre site (Site) at 21730 Newland Street in the City of Huntington Beach, California. The Site location is shown in Figure 2-1. Also appearing in Figure 2-1 is the AES Corporation (AES) Huntington Beach Generating Station (HBGS), which is located adjacent to the Site.

In order for the Project to proceed, a Coastal Development Permit must be obtained from the California Coastal Commission (CCC). A permit application was submitted, and CCC Staff [July 13, 2012] responded with a Notice of Incomplete Coastal Development Permit Application.

The Notice of Incomplete Coastal Development Permit Application [CCC Staff, July 13, 2012] requested additional information in two main categories: Geologic Hazards and Project Stability; and Tsunami Hazards and Risks. The Geologic Hazards and Project Stability-related request included:

• Surface fault rupture and structural stability;

• Lateral soil spread;

• “Design-level” earthquake; and

• Liquefaction.

The Tsunami Hazards and Risks-related request included:

• Site-specific information;

• Use of updated studies and information; and

• Design modifications and mitigation measures.


2.2 Geosyntec Scope-of-Work

The work conducted by Geosyntec addresses the CCC Staff’s request regarding results of the required investigations, studies, and analyses identified in the project’s Subsequent Environmental Impact Report (“SEIR”) Mitigation Measures GEO-l through GEO-9.

The purpose of this Report is to address the issues raised by CCC Staff [July 13, 2012]. In order to address these items, Geosyntec performed the following scope of work:

• Review of available existing information for the Site;

• Supplemental geotechnical field investigation;

• Geotechnical site characterization;

• Development of a geotechnical design basis;

• Seismic hazard analysis and development of design ground motions;

• Site response analysis;

• Surface fault rupture hazard evaluation;

• Liquefaction and lateral spreading analysis;

• Tsunami hazard analysis; and

• Development of recommendations.


3. FIELD INVESTIGATION

3.1 Field Investigation

The geotechnical field investigation consisted of 5 Cone Penetration Test (CPT) soundings ranging in depth from approximately 50 ft to approximately 98 ft. The CPT soundings were performed on 18 October 2012 under Geosyntec supervision using a 30-ton CPT rig.

In total, 5 CPT soundings were performed. Locations of these 5 soundings are shown in Figure 2-1 and designated as CPT-01 through CPT-05. The figure shows these CPT sounding with respect to Site topography, locations of existing improvements, and locations of the proposed facilities.

The soundings were advanced to depths of approximately 96 to 97 ft below ground surface (ft bgs) for CPT-01 through CPT-03, and approximately 50 ft bgs for CPT-04 and CPT-05. In addition, shear wave velocity (Vs) tests were conducted at 10-ft intervals for soundings CPT-01 through CPT-03 to assist in developing material properties for seismic site response analysis.

Appendix A-1 contains the CPT data recorded during the field investigation. Interpretation of the CPT data and information from previous investigations is included in Section 4 of this report.

3.2 Review of Previous Field Investigations

Geosyntec also reviewed geotechnical reports prepared by GLA [2002], Magorien [2002a], Magorien [2002b] and Magorien [2010] for the Site and by Ninyo & Moore [2011] for the neighboring AES HBGS. Data derived from the GLA [2002] and Ninyo & Moore [2011] investigations are enclosed in Appendices A-2 and A-3, respectively.

GLA [2002] presents the results of a total of 21 CPT soundings4 and 6 hollow-stem auger borings. Locations of GLA [2002] CPT soundings and hollow-stem auger borings are shown in Figure 2-1. These CPT soundings and borings were advanced to depths ranging from approximately 82 ft bgs to approximately 93 ft bgs.

4 From the GLA field investigation, CPT-21A, is located outside of the Site boundary.


Magorien [2002a], [2002b], and [2010], each give a summary of the regional settings, subsurface conditions and geotechnical constraints. Magorien [2002a], [2002b], and [2010], provide an assessment of the GLA [2002] report along with a summary of the seismicity, liquefaction and lateral spreading and the impacts and mitigation measures to consider. No field investigation data are included in these reports.

Ninyo & Moore [2011] advanced a total of 2 hollow-stem auger borings and 4 CPT soundings5. The borings were advanced to a depth of approximately 51.5 ft bgs. The CPT soundings were advanced to depths ranging from approximately 54 ft to approximately 75.5 ft bgs.

5 From the Ninyo and Moore field investigation, one boring and one CPT sounding are located within the Site boundary.


4. SUBSURFACE CONDITIONS

4.1 General Geologic Conditions

Figures 4-1 and 4-2, respectively, show a geologic map and a geologic cross section for the Site vicinity. As these figures indicate, the Site is generally underlain by alluvial deposits consisting of a combination of sands, gravels, silts, and clays. Based on the geologic cross section shown in Figure 4-2, which is taken from the Orange County Water District’s Groundwater Management Plan [OCWD, 2004], the depth to bedrock is understood to be approximately 200 ft to 300 ft bgs.

4.2 Subsurface Soil Conditions

Using the information from the current CPT soundings and previous investigations by GLA [2002] (Appendix A-2) and Ninyo & Moore [2011] (Appendix A-3), Geosyntec developed an idealized soil profile (Figure 4-3) for the Site. In general, subsurface soils from ground surface to a depth of approximately 98 ft bgs were found to consist of:

• Approximately 10 ft of compacted sandy clay fill; underlain by

• Approximately 2 ft of high-plasticity clay; underlain by

• Approximately 38 ft of fine sand to silty fine sand; underlain by

• Approximately 20 ft of well-graded sand; underlain by

• Alternating layers of sandy clay and silty sand.

Bedrock was not encountered within the depths explored in either the current or previous investigations, and is instead understood as described in Section 4.1.

In addition to Site stratigraphy, Figure 4-3 presents material properties estimated for the soil strata listed above. These properties were developed based on review of available laboratory data, correlations with CPT measurements, Geosyntec’s experience at nearby sites, and typical values from the literature. The idealized profile and properties in Figure 4-3 were used in subsequent analyses, as described below.


4.3 Groundwater Conditions

In discussing the water level at the Site, it is important to recognize that the North American Vertical Datum of 1988 (NAVD 88) mean sea level is distinct from the local mean sea level (i.e., the mean sea level elevation in the vicinity of the Site).

Where the phrase mean sea level is used in this Report, the term refers to the NAVD 88 elevation datum. Elevations cited within this Report are based on this NAVD 88 mean sea level datum. NAVD 88 is the current standard vertical control survey point in the United States of America based upon the General Adjustment of the North American Datum of 1988.

Where the qualifier local is used, as in local mean sea level, the term refers to the water elevation at the shoreline near the Site. Local mean sea level was assumed to have an elevation of +2.6 ft above mean sea level (ft msl) based on data for the Los Angeles monitoring station available on the National Oceanic and Atmospheric Administration (NOAA) website (www.noaa.gov).

Tetra Tech [2012] reported groundwater encountered at depths ranging from 6 to 10 ft below ground surface (ft bgs)6, and indicated that groundwater elevation at the Site is tidally influenced. Although ground surface elevation varies across the Site, a value of approximately +10 ft msl is considered representative. Assuming a representative ground surface elevation of +10 ft msl and a representative depth to groundwater of approximately 8 ft bgs, a representative groundwater elevation at the Site is estimated to be approximately +2 ft msl. This value is approximately consistent with the estimated local mean sea level of +2.6 ft msl. Using this information coupled with the NOAA monitoring station data, the local mean high water level near the Site is estimated to be approximately +4.6 ft msl.

6 One recording in the Tetra Tech memorandum (CPT-3) has a depth to water of 16.5 feet bgs and is thought to be an outlier as it is 8-10 feet deeper than in the surrounding CPTs.


5. DESIGN BASIS

In order to provide a basis for the recommendations stated in Section 12, it is necessary to develop a set of design criteria for the Site. Calculated values (e.g., factor of safety against liquefaction triggering) are compared against Site baseline design criteria. In general, where calculated values for the Site fail to meet the baseline design criteria, mitigation is recommended. In general, where calculated values for the Site meet or exceed the design criteria, further mitigation is considered to be unnecessary.

The design basis for this site has been established in accordance with the requirements of the 2010 Edition of the California Code of Regulations [CCR, 2010], local practice, and the anticipated 30-year design life of the project. In particular, the following codes and practice guidance documents have been considered:

• the 2009 Edition of the International Building Code [IBC, 2009];

• Idriss and Boulanger [2008];

• ASCE 7-05;

• SP 117 [CDMG, 1997];

• SCEC [1998]; and

• LA DPW [2000].

The IBC [2009] is the basis for CBC [2010]. ASCE 7-05 is referenced in IBC [2009]. The CDMG [1997] is a document that defines standard practice for evaluation and mitigation of seismic hazards in California. Idriss and Boulanger [2008] is considered the state of the practice guideline for liquefaction susceptibility and triggering. SCEC [1998] is a guidance document for implementation of CDMG [1997] related to soil liquefaction. The County of Los Angeles Department of Public Works [LA DPW, 2000] is a guidance document for preparation of Geotechnical Reports in Southern California.

Construction is common in high seismicity locations, especially along the California Coast where seismic loading criteria may exceed the baseline design criteria listed above. In such cases, the implementation of enhanced design and construction measures can provide sufficient mitigation to protect facility structural integrity.


6. SEISMIC HAZARD ANALYSIS

6.1 Seismic Hazard Analysis in support of Design

The Site is in an area of high seismicity (common for much of Southern California). The main trace of the Newport-Inglewood Fault is within approximately one half mile (0.5 mile) northeast from the Site. Other significant faults potentially affecting the Site include the Compton Blind Thrust, Palos Verdes Fault, Elysian Park Blind Thrust, Whittier Fault, and several other faults at a distance greater than 20 miles from the site. With a Moment Magnitude Mw 7.17 assigned by the United States Geological Survey (USGS; see Petersen et al., 2008)8 and a relatively short site-to-source distance of 0.5 miles (0.8 km), the Newport Inglewood Fault is the governing seismic source for this site.

To further assess the seismic hazard at the site, in accordance with the CBC [2010] requirements, Geosyntec performed a probabilistic seismic hazard analysis. In accordance with the CBC [2010] requirements, the design basis is an event with 2 percent probability of exceedance in 10 years (2% PE in 10 years). The return period of this design event is 2,475 years.

The probabilistic seismic hazard analysis was performed by means of the 2009 deaggregation tool and the 2003 Java Seismic Hazard Calculator (latest update), as posted on the USGS web site. The analysis was performed for the geometric center of the site (33.6473 degrees North Latitude and -117.9774 degrees West Longitude). The results are presented in Figures 6-1 and 6-2.

Figure 6-1 presents the deaggregated seismic hazard for the site. This Figure indicates that the deaggregated Mw (Modal) is 7.02. The corresponding bedrock Peak Horizontal Ground Acceleration (PHGA) is 0.61 g. Geosyntec notes that the deaggregated Mw of 7.02 is slightly lower than the commonly cited Mw for the Newport Inglewood Fault of 7.1, as the former is based upon the 2008 information and interpolation between grid lines, as incorporated in the USGS database, while the latter is fault-specific and is up to date. Therefore, to provide a conservative basis for design the later value of Mw 7.1 is used for design.

7 See discussion in Section 6.2 of this report. 8 http://pubs.usgs.gov/of/2008/1128/


Figure 6-2 shows the 5% viscous damping acceleration response spectrum for free-field bedrock (i.e., IBC 2009 Site Class B) conditions at the site. This acceleration response spectrum serves as a basis for development of bedrock design ground motions at the site, as explained in Section 7.

6.2 Discussion on the Magnitude assigned to the Newport-Inglewood Fault

As explained above, the Mw for the Newport Inglewood fault obtained by deaggregation of seismic hazard using the latest (i.e., 2009) version of USGS web tool is Mw 7.02 (modal value; results of deaggregation on Figure 6-1).

The USGS probabilistic seismic hazard analysis which produced this value (i.e., deaggregated Mw 7.02) considers multiple scenarios (i.e., sets of fault parameters, including ranges of Moment Magnitudes) for the Newport Inglewood fault. These scenarios are assigned labels (e.g., Scenario 1, Scenario 2, etc.) for identification. Moment Magnitude values in these scenarios include:

• Mw 6.5 (lower bound of Mw range; Scenarios 1 and 2)9;

• Mw 7.1 (upper bound of Mw range; Scenario 1)10; and

• Mw 7.5 (upper bound of Mw range; Scenario 2)11.

Each scenario is assigned a weight. The probabilistic seismic hazard analysis considers each scenario in proportion to its assigned weight and produces a representative Moment Magnitude value.

6.3 Supplemental Seismic Hazard Analysis – Parametric Study

During the technical exchange meeting between CCC Staff and Geosyntec on February 27, 2013, CCC Staff expressed the following concerns: (i) that the Moment Magnitude associated with the Newport-Inglewood fault may be as high as Mw 7.5; and (ii) that the site-to-source distance might be less than 0.5 miles (0.8 km; as evaluated based upon

9 http://geohazards.usgs.gov/cfusion/hazfaults_search/disp_hf_info.cfm?cfault_id=127ab 10 http://geohazards.usgs.gov/cfusion/hazfaults_search/disp_hf_info.cfm?cfault_id=127ab 11 http://geohazards.usgs.gov/cfusion/hazfaults_search/disp_hf_info.cfm?cfault_id=127_alt2


USGS data imported in Google Earth). Geosyntec performed a parametric study to address these CCC Staff concerns.

6.3.1 Moment Magnitude for the Newport-Inglewood Fault

Following the approach of probabilistic seismic hazard analysis is the standard practice for evaluating a site’s seismic hazard. This approach uses fault parameters from a centralized USGS database. Although the USGS database may be updated from time to time, these parameters are generally considered to be fixed.

As mentioned above in Section 6.2, the probabilistic seismic hazard analysis performed for the Site produced a magnitude of Mw 7.02. As this value relies on USGS parameters, and USGS parameters are considered to be fixed, the magnitude from the probabilistic seismic hazard analysis approach (i.e., Mw 7.02) is considered to be fixed, also. Therefore, a different approach is required to consider a magnitude of Mw 7.5 for the Newport-Inglewood fault.

To address CCC Staff concerns, Geosyntec followed a deterministic seismic hazard analysis approach, assigning a magnitude of Mw 7.5 to the Newport-Inglewood fault. The analysis was performed using a suite of 5 New Ground Attenuation (NGA) attenuation relationships that are commonly abbreviated as A & S, B & A, C & B, C & Y and Idriss (see, e.g., Abrahamson et al., 2008). These attenuation relationships are in wide use for shallow crustal earthquakes and are also incorporated in the USGS web tools.

6.3.2 Site-to-Source Distance

To address the CCC Staff comment regarding site-to-source distance, Geosyntec varied the site-to-source distance used in the deterministic seismic hazard analysis. Site-to-source distances considered ranged between 0.01 and 100 km.

6.3.3 Parametric Study Results

The results of Geosyntec’s deterministic seismic hazard analysis are shown in Figure 6-3. Figure 6-3 indicates that when a fault capable of generating Mw 7.5 is placed at a site-to-source distance of 0.01 km (i.e., conservatively assumed to be below the Site), the calculated bedrock PHGA is approximately 0.58 g.


This value of 0.58 g is lower than the value of 0.61 g calculated from the probabilistic seismic hazard analysis (see Section 6.1 of this Report). Therefore, the results of the probabilistic seismic hazard analysis documented herein are considered to be conservative and are recommended to be used for engineering evaluations in support of design of the Poseidon Desalination Plant.


7. EVALUATION OF DESIGN GROUND MOTIONS

The design ground motions (i.e., accelerograms; acceleration time histories) are an essential input to a site-specific, site response analysis.

A suite of acceleration time histories representative of the design earthquake was selected in accordance with the ASCE 7-05 requirements. Geosyntec selected a suite of five acceleration time histories that envelope the target (i.e., bedrock) acceleration response spectrum. The following methodology was used: (i) screen the database of acceleration time histories on the basis of earthquake magnitude to select a reduced set of accelerograms that were recorded in events similar to the design earthquake with respect to magnitude and PHGA; and (ii) plot the acceleration response spectra of the candidate accelerograms against the target acceleration response spectrum and select the representative accelerogram for use in the design analyses.

By using the above methodology, Geosyntec selected the following five accelerograms to represent design (i.e., bedrock) ground motions at the site:

• The Big Bear Lake accelerogram recorded during the Mw 6.7 Big Bear, California earthquake;

• The 360-degree component of the Castaic – Old Ridge Route Accelerogram from the Mw 6.7 Northridge earthquake (Ds = 9.06 seconds);

• The North-South (N-S) component of the Olympia accelerogram from the 1949 Mw 7.1 Western Washington earthquake (Ds = 18.8 seconds), scaled to 0.57 g;

• The 90-degree component of the Silent Valley accelerogram from the Mw 6.9 Loma Prieta earthquake; and

• The S69E component of the Taft - Lincoln School Tunnel record from the 1952 Mw 7.4 Kern County earthquake (Ds = 30.6 seconds), scaled to 0.36 g.

The acceleration response spectra of the selected time histories are plotted against the target acceleration response spectrum on Figure 7-1. Figure 7-1 indicates that the acceleration response spectra of these accelerograms match the target spectrum over the range of periods considered (0.2 to 4 seconds). Therefore, these accelerograms, scaled to PHGA of 0.61 g, were selected as the representative acceleration time histories for use in seismic site response analyses at the Site.


8. SITE RESPONSE ANALYSIS

Geosyntec conducted a site response analysis to evaluate the influence of the local soil conditions (i.e., of local terrace deposits) on the design (i.e., bedrock) motions. Given the relatively flat site topography and horizontally layered ground conditions, a one-dimensional (1-D) site response analysis was considered appropriate. Following the recommendations in Ishihara [1986] for sites expected to experience relatively high bedrock PHGA, a non-linear site response analysis was conducted. Additionally, because the Site is considered to be vulnerable to liquefaction, the Site is considered a Site Class F per the CBC [2010], and therefore required a site-specific response analysis. This section provides the required site response analysis.

An idealized soil profile for the site, developed on the basis of site-specific investigations documented in Section 4, is shown in Figure 4-3. This profile indicates that on-site materials to a depth of approximately 97 ft, can be divided into the following groups: (i) artificial fill; (ii) marine clays; and (iii) silty sands.

Material parameters assigned by Geosyntec to the above-listed materials were derived from the results of shear wave velocity measurements documented in Appendix A-1. Unit weights are assumed consistent with soil liquefaction analysis (see Section 10 of this report). A shear wave velocity value of 2,500 ft/s (731 m/s) was conservatively assigned to a hard pan encountered by CPT soundings at an approximate depth of 97 ft bgs based upon Burger [1992]. The Vucetic and Dobry [1991] modulus reduction and damping curves assigned to the materials in the profile are shown in Figure 8-1 (PI = 0, 30 and 50 were used).

One-dimensional non-linear seismic site response analyses of the site were conducted using the computer program D-MOD2000 [Matasovic, 1993; Matasovic and Vucetic, 1995; www.GeoMotions.com]. D-MOD2000 solves the dynamic equation of motion in the time domain. The behavior of soil and soil-like materials is modeled in D-MOD2000 using the non-linear hysteretic constitutive model proposed by Matasovic and Vucetic [1993]. A small amount of viscous damping (typically in the range of 0.5 to 1.0 percent) is assigned to the soil materials in the profile to ensure the numerical stability of the solution.

Parameters for the Matasovic and Vucetic [1993] constitutive model were developed by curve-fitting the above-cited modulus reduction and damping curves. The five


representative accelerograms, scaled to the design bedrock PHGA of 0.61 g, were applied to the base of the 1-D site response analysis column. The representative column is approximately 97 ft high, consistent with the idealized soil profile for the site.

A summary of the material input data for site response analysis, as input in D-MOD2000, is attached as Appendix B. D-MOD2000 outputs are also enclosed in Appendix B. The results of the site response analyses are summarized in Figures 8-2 and 8-3.

Figure 8-2 presents the results in terms of calculated free-field (ground surface; no structure) elastic acceleration response spectra (5 percent viscous damping). Both spectra of individual accelerograms and an average spectrum of all calculated surface accelerograms are shown, as required by ASCE 7-05. Figure 8-2 indicates that, over the period range of interest, the approximately 97-ft deep soil profile (i.e., 1-D column) attenuates the input (i.e., bedrock) motions.

Per CBC [2010] requirements, the average acceleration response spectrum calculated from the site-specific response analysis should be compared against a code-based minimum. This code-based minimum spectrum corresponds to 80% of the probabilistically-established acceleration response spectrum for a Site Class E site.

Figure 8-2 shows this code-based minimum spectrum (i.e., black dashed line) with respect to the average spectrum calculated from the site response analysis (i.e., solid black line). Because the spectrum calculated from the site response analysis is below the code-based minimum spectrum, CBC [2010] requires that the 80% Site Class E spectrum (i.e., black dashed line) should be used for design at this Site. Based on the “design-level” earthquake analysis, the following design measure is recommended:

• Design Measure A: Geosyntec recommends that the Project be designed following CBC [2010] requirements for Site Class F, with an acceleration response spectrum corresponding to 80% of the Site Class E response spectrum.

Figure 8-3 shows the calculated peak shear stress profile. The average value of this peak shear stress profile is an essential input for evaluation of soil liquefaction profile at this site.


9. SURFACE FAULT RUPTURE HAZARD

According to the draft City of Huntington Beach [2011] Local Hazard Mitigation Plan and the City of Huntington Beach [1996] General Plan, the Site may be traversed by a fault referred to as the South Branch Fault.

According to the Southern California Earthquake Center (SCEC, http://www.data.scec.org/significant/newport.html), the Newport-Inglewood fault system did not create a surface rupture during the 1933 Long Beach earthquake (Mw 6.4). Additionally, the South Branch Fault is not mapped by USGS as a Holocene fault. Furthermore, as evidence that the South Branch Fault actually passes below the Site is limited, this fault is referred to here as a postulated fault.

Nevertheless, the CCC Staff [July 13, 2012] has expressed concern that a subsurface rupture of this postulated fault may propagate to the surface and damage the proposed facilities. The following discusses the likelihood of a surface fault rupture scenario and presents an engineering evaluation of possible surface effects of subsurface faulting below the Site.

9.1 Likelihood of Surface Fault Rupture at the Site

The South Branch Fault is a branch fault of the Newport-Inglewood fault system. Portions of this fault system are recognized to be active, and fault ruptures within this system are recognized as having caused several recorded seismic events, including the 1920 Inglewood earthquake and the 1933 Long Beach earthquake.

Several researchers, including Wood [1933], Barrows [1974], and Yeats et al. [1981] have studied this fault system and the surface effects of seismic events originating from it. According to Wood [1933], who surveyed the aftermath of the 1933 Long Beach earthquake, “No fresh movement of faulting extending to the surface has been observed anywhere on this occasion.” Regarding the Newport-Inglewood fault in general, Barrows [1974] stated, “No surface faulting along known faults has been observed resulting from historic earthquakes.” Yeats et al. [1981] characterized the Newport-Inglewood fault system as having “…a demonstrated seismic-shaking potential, but it has not shown evidence of ground rupture…”

Based on these reported observations, it appears that the Newport-Inglewood fault system, in general, has a limited potential for surface fault rupture impacts. Because the


South Branch Fault is a part of this fault system, it is reasonable to assume that this conclusion also applies to the South Branch Fault. As the South Branch Fault appears to be the only postulated fault traversing the Site, the possibility of a surface fault rupture at the Site is considered to be unlikely.

9.2 Surface Fault Rupture Finite Element (FE) Analysis

While the exact location of the South Branch Fault and its proximity to the Site is unknown, for purposes of this analysis it was conservatively assumed that the fault traverses the Site. Although fault rupture propagation to the surface of the Site is considered to be unlikely, as explained in the previous section, possible surface impact of subsurface fault rupture was conservatively evaluated using a Finite Element (FE) modeling approach. An FE model was developed using SIGMA/W software (www.geo-slope.com) for one representative cross section (i.e., cross section A-A’) through the Site.

9.2.1 Model Geometry

Figure 2-1 shows the location of cross section A-A’ in plan view. Figure 9-1 shows this cross section in profile. The FE model geometry, shown in Figure 9-2, was developed based on the cross section shown in Figure 9-1.

While the proposed building structures were assumed to have a maximum height of 35 ft, the model results are relatively insensitive to building heights less than 3 stories. Although it is considered likely that ground improvement (e.g., stone columns) or deep foundation systems (e.g., piles) will be used at the Site, for the purposes of this analysis, the proposed buildings were conservatively assumed to be constructed on concrete mat foundations. The foundations were assumed to be 2 ft in thickness and to be placed on the ground surface. It should be re-emphasized that the shallow foundation assumption is conservative and in general, deep foundation systems like piles could improve the performance of the structure.

A soil zone was estimated to lie between the ground surface and the estimated location of the top of bedrock. The elevation of the top of bedrock was estimated as -200 ft above mean sea level (ft msl) based on the geologic cross section shown in Figure 4-2. The groundwater table was estimated to be approximately the same as the local mean sea level. This local mean sea level was estimated to have an elevation of 2.6 ft above


mean sea level (ft msl) based on data for the Los Angeles monitoring station available on the NOAA website (www.noaa.gov), as stated in Section 4.3.

The location of the South Branch Fault was assumed based on its postulated location shown on the fault map in the City of Huntington Beach Draft Hazard Mitigation Plan [City of Huntington Beach, 2011]. The assumed location of the fault is shown in Figures 9-1 and 9-2. The South Branch Fault was estimated to act as a normal fault (i.e., the hanging wall was estimated to move downward and away relative to the footwall). In order to simulate rupture of the South Branch Fault, a prescribed displacement boundary condition was applied at the location of the fault along the assumed top of the bedrock.

9.2.2 Material Properties

The foundation was modeled as a linear elastic material. The foundation was assigned a Young’s modulus of E=0.6x109 psf and a Poisson’s ratio of ν=0.15.

The soil was modeled as a linear elastic-perfectly plastic Mohr-Coulomb material. The soil Poisson’s ratio was assumed to have a constant value of ν=0.3, but the soil Young’s modulus was assumed to vary with depth. Values for soil Young’s modulus in the top 97 ft bgs were taken as one-half of the small-strain Young’s modulus estimated from seismic CPT results. Soil Young’s modulus values below 97 ft bgs were estimated by extrapolation using a power law function. The soil was assumed to have a friction angle of φ=35°.

9.2.3 Magnitude of Model Fault Displacement

In order to estimate the magnitude of the subsurface fault displacement at the Site, the following key assumptions were made:

• The South Branch Fault is a secondary fault within the Newport-Inglewood fault system;

• The main trace of the Newport-Inglewood fault system lies approximately one-half mile from the South Branch Fault; and

• Displacement on the South Branch Fault occurs during a moment magnitude M=7.1 event on the main trace of the Newport-Inglewood fault system.


Magnitude of the displacement on the main trace was estimated according to the empirical relationship between moment magnitude and surface displacement proposed by Wells and Coppersmith [1994], as shown in Figure 9-3. For the assumed moment magnitude, average surface displacement was estimated to be approximately 3.8 ft on the main trace of the fault.

The magnitude of the displacement on the South Branch Fault was estimated according to the measurements of secondary fault displacement as a percentage of total fault displacement reported by Lazarte [1996], as shown in Figure 9-4. Based on these measured values, a value of 25% of total fault displacement (i.e., 25% of 3.8 ft, or approximately 0.95 ft) was assumed for the magnitude of the South Branch Fault displacement.

9.2.4 Structural Damage Thresholds

In order to assess the degree of structural damage likely to result from fault rupture below the Site, it is necessary to establish a set of structural damage criteria. A set of criteria based on angular distortion (i.e., the difference in vertical displacement between two points divided by the distance between those two points) is considered to be appropriate for the model geometry considered here. Salgado [2008] recommends a set of angular distortion-based criteria for varying levels of structural damage. The levels of structural damage considered by Salgado [2008] are Serviceability Limit State (SLS, i.e., architectural damage: wall cracking is likely to be initiated) and Ultimate Limit State (ULS, i.e., structural damage: frame cracking is likely to be initiated). According to Salgado [2008], SLS and ULS are likely to occur at the following angular distortion thresholds:

• 1/500: unlikely to lead to either SLS or ULS;

• 1/300: SLS (wall cracking); and

• 1/170: ULS (frame cracking).

9.2.5 Model Results

Figure 9-5 shows the relevant results of the FE analysis. The upper portion of the figure shows calculated vertical displacement contours and deformed mesh for the region of the model near the location of the proposed structures. The lower portion of the figure


shows model-calculated angular distortion along the bottom of the foundation for both of the proposed structures included in the analysis.

As shown in Figure 9-5, the maximum model-calculated angular distortion is approximately 1/276 near approximate station 440 ft on the cross section analyzed. This model-calculated value is more severe than the SLS of 1/300 but less severe than the ULS of 1/170.

9.3 Fault Rupture Conclusion

The fault rupture results suggest that, for the fault rupture scenario analyzed, the proposed structures may experience repairable aesthetic and temporary serviceability issues, but significant structural damage is considered to be unlikely. Based on results of the investigation and analysis, no changes to the project layout, design, engineering or mitigation measures are needed to avoid fault-rupture related hazards to structural stability.


10. LIQUEFACTION AND LATERAL SPREAD HAZARD ASSESSMENT

The term liquefaction refers to a sudden loss of soil strength due to pore pressure buildup in response to a loading event such as earthquake shaking. Experiences from previous earthquakes have demonstrated that loose granular soils located near the ground surface and saturated by a high water table are the most susceptible to liquefaction. A related phenomenon is lateral spreading where liquefied soil located near a free face or on gently sloping ground, such as near the canal, moves as a mass, resulting in deformations and application of lateral forces on structures and their foundations.

Given the significant deposits of sands beneath the project Site, there is potential for liquefaction during an earthquake, with the resulting consequences including both settlement and lateral spreading towards the adjacent Huntington Beach flood control channel. This section presents Geosyntec’s evaluation of liquefaction potential (triggering) and consequences based on the five CPTs completed during the current investigation. Liquefaction triggering analyses were previously performed on historic CPTs at the site, and as the observed behavior between historic and current CPTs were similar, the current CPTs performed under Geosyntec’s supervision were considered suitable for demonstrating the anticipated behavior at the Site. Figure 10-1 shows the locations of the CPT soundings along with the stability cross-sections utilized for evaluation of the lateral spread hazard.

10.1 Liquefaction Triggering

The soils within each of the five recent CPTs on site (CPT-01 through CPT-05) were analyzed for liquefaction triggering. Evaluations were performed for the design seismic event corresponding to a Mw 7.1 earthquake on the Newport Inglewood fault, with a free-field peak ground acceleration (PGA) of 0.33g based on the site response analysis presented in Section 8 (see Figure 8-2).

All of the CPTs advanced on site (Figure 10-1) encountered relatively loose, clean sands that show susceptibility to liquefaction at the design levels of shaking according to several methods, including the Youd et al. [2001] procedure, Robertson [2010], and Idriss and Boulanger [2008]. The Idriss and Boulanger [2008] procedure, which tended to predict somewhat more liquefaction at depth than the other two methods, was used to develop the liquefaction triggering profiles presented in Figure 10-2. These profiles


present factors of safety (FS) against liquefaction computed at discrete depths in each of the five recent CPTs on site. Per Idriss and Boulanger [2008], liquefaction susceptibility is generally defined as a factor of safety less than 1.2.

Zones within the CPTs with susceptibility to liquefaction are shown on the “post-liquefaction” idealized soil profile on Figure 10-3. Two liquefiable zones were identified in the subsurface soils: (i) an upper layer approximately 4 ft-thick; and (ii) multiple lenses between 45 and 70 ft bgs. The presence of these potentially liquefiable soil lenses below 45 ft bgs was also observed in Geosyntec’s review of CPTs from previous investigations to similar depths (e.g., GLA [2002] and Ninyo and Moore [2011]).

10.2 Effects of Liquefaction

Liquefaction involves soil strength reduction and may lead to vertical and/or lateral ground deformations. The following discussion presents effects of liquefaction, including residual soil strength upon liquefaction triggering, anticipated reconsolidation settlement, and potential lateral spread deformations.

10.2.1 Residual Shear Strength of Liquefied Soil

The shear strength of liquefied soil is referred to here as the soil’s residual shear strength. This value is generally evaluated based on empirical methods for estimation

of a ratio of residual shear strength to vertical effective stress (su(LIQ)/σ’v). This ratio provides for the anticipated increase in residual shear strength with increasing depth in a given soil profile. Figure 10-4 shows the residual shear strength analysis for all CPTs based on the Idriss and Boulanger [2008] method. Residual shear strength was estimated for soil strata with a factor of safety against liquefaction lower than 1.2 (Figure 10-2). A residual shear strength ratio of 0.05 in the upper liquefiable layer and 0.1 in the lower liquefiable zone was selected based on review of the scatter in the results (Figure 10-4). These results were subsequently used in the evaluation of lateral spread potential.

10.2.2 Liquefaction Induced Settlement

Post-liquefaction vertical settlement generally occurs when the excess pore pressures generated during the earthquake dissipate and the loose sands reconsolidate.


Reconsolidation settlement is estimated here following the strain potential approach detailed by Ishihara and Yoshimine [1992]. For this approach, reconsolidation-induced vertical strain is estimated at multiple points within the liquefiable zones identified. This estimated vertical strain is then integrated over the thickness of the liquefiable zones to estimate total reconsolidation settlement.

Each of the five locations investigated by Geosyntec was evaluated using this approach. Results of the evaluation are shown in Figure 10-5 and summarized in Table 10-1. Figure 10-5 shows both incremental estimated settlements in each liquefiable zone with FS less than 1.2 (Figures 10-2 and 10-3), and cumulative settlements developed by adding the incremental settlements over the total length of each CPT sounding. Based on these evaluations, it is estimated that up to 9 in. of total reconsolidation settlement may occur at the Site. It is recommended that the Project’s Structural and Geotechnical Engineers collaborate on the design of the foundation system to accommodate the anticipated settlement. Several geotechnical mitigation methods that can reduce settlements to acceptable levels are described briefly in Section 10.3.

10.2.3 Evaluation of Lateral Spreading Displacements

The term lateral spreading refers to the movement of a soil mass down-slope or toward a free face as a result of liquefaction of subsurface soils. Lateral spreading may result in significant ground deformations, vertical settlements and additional lateral forces on structures.

Several approaches for estimating lateral spreading displacements exist. These include approaches related to shear strain potential (e.g., Zhang et al., 2004), empirical relationships based on case histories (e.g. Youd et al., 2002), and methods based on the Newmark sliding block approach (e.g. Bray and Travasarou, 2007). Two approaches were selected for this evaluation, namely, the Zhang et al. [2004] and Bray and Travasarou [2007] methods. The Youd et al. [2002] method is not well calibrated for faults in close proximity to the site, such as the case at hand. Note that given the complexity of the lateral spreading phenomenon, all methods for computing lateral displacement are considered approximations.


Strain Potential Approach

The strain potential approach for estimating lateral spread deformations [Zhang et al., 2004] involves estimating permanent strains within liquefied zones and integrating over the thickness of the zone to compute a total estimated lateral displacement index (LDI). The LDI is then used as an input to estimate the lateral spreading displacement based on site conditions (e.g., height of the free face, distance to the free face, etc.).

Lateral spreading displacement was estimated based on information from Geosyntec’s five CPT soundings. Results of the evaluation are presented in Figure 10-6 and summarized in Table 10-2. The estimations predict LDIs ranging from 30 inches to 44 inches. Note that LDI is an index later used in the prediction of displacement and, as described subsequently, lateral spread displacements will be lower than these values depending on site conditions.

Geosyntec conservatively estimated the LDI considering liquefiable soils within the first 50 ft bgs. A free face height of 13 ft was assumed based on the approximate depth of the Huntington Beach Flood Control Channel shown on the Base Map of Drainage Facilities in Orange County (www.ocflood.com). Chu et al. [2006] compared measured vs. predicted lateral spread displacements from the Chi-Chi, Taiwan earthquake and found better agreement between the two when they limited the depth of application for LDI to 2 x H, where H is the height of the free face (i.e. 2 x H = 26 ft). Idriss and Boulanger [2008] recommend caution when using this limit where significant liquefiable material exists below this depth. Increasing the depth of influence to 50 ft captures the upper portion of the lower liquefiable zone, and provides a reasonable lower bound based on the geometry of the canal, which is on the order of 70 to 80 ft wide. Geosyntec considers that liquefaction occurring more than 50 ft from the ground surface is unlikely to influence a lateral spread into this relatively narrow canal. Slope stability analyses presented below confirm that this is a reasonable assumption. The distance from proposed Site structures to the free face was assumed to be approximately 150 ft (closest approach of the channel to a site structure at the northeast corner). Table 10-2 lists the lateral displacements estimated at approximately 150 ft from the free face. At this distance, the estimated lateral spreading displacement on the Site ranges from approximately 26 in. to 38 in. toward the Huntington Beach Flood Control Channel. Note that lateral spreading displacements would be expected to decrease at greater distances from the channel.


Newmark Sliding Block Approach

The Newmark sliding block approach applies an earthquake acceleration time history to a critical sliding mass and computes lateral displacements when the applied accelerations exceed a yield acceleration (ky) that results in a factor of safety against sliding equal to unity. The yield acceleration, computed using pseudo-static slope stability analyses, along with earthquake ground motion characteristics, can be used in regression relationships to estimate seismically induced displacements. Bray and Travasarou [2007] present a detailed regression analysis of a set of case histories to develop the relationships used herein.

Three representative cross sections were chosen to evaluate lateral spread displacement using the Newmark sliding block approach. These sections are SA-SA’, SB-SB’, and SC-SC’ as located on Figure 10-1 and shown in profile on Figures 10-7, 10-8 and 10-9. Aside from modest topography changes, the primary difference between the sections is the geometry of the canal. The Base Map of Drainage Facilities in Orange County (available at www.ocflood.com) shows the location and approximate dimensions of the canal and depth of sheet piles that form the canal on the northern side and northeastern corner of the site. The canal along much of the western margin of the site is formed by a sloping embankment without a sheet pile. The canal is identified as 13 ft deep and 70 to 80 ft wide. The sheet piles are shown to have a total length of 34 ft (i.e. 21 ft of embedment).

Stability analyses were performed using the program Slope/W [Geo-Slope International, 2007] using Spencer’s method [Spencer, 1967]. Trial potential failure surfaces are generated based on a grid of circle centers and a series of tangent lines. Both circular and non-circular stability surfaces can be evaluated. Non-circular surfaces are used when there are specific material layers that may control the shape of a slope failure. Several cases were evaluated with non-circular surfaces to simulate a stability surface that passes preferentially through a weak liquefied zone. This condition can be controlled in Slope/W by defining an impenetrable layer (“bedrock”) below the target liquefiable zone.

An initial attempt was made to model the sheet pile wall within Slope/W. However, while the software does have the capability to incorporate a sheet pile element, it is not


well suited to seismic conditions. In addition, none of the methods for evaluating lateral spread deformations can account for the influence of a sheet pile wall. As such, while the sheet pile wall was included in static stability analyses, the pseudo-static analyses focused on evaluations of shallow slip surfaces that ignore the presence of the wall, and deep slip surfaces that extend below the sheet pile wall and exit within the canal. These scenarios were used to evaluate the range of anticipated behavior at the site rather than predict the influence of the sheet pile wall.

Two scenarios were considered for the stability analyses:

1. Shallow Pseudo-Static Stability – residual strength applied to upper liquefiable zone per Figure 10-3 and surface forced to fail through liquefied soil by defining underlying soils as “bedrock”;

2. Deep Pseudo-Static Stability – residual strength applied to upper and lower liquefiable zones per Figure 10-3 and surface forced to fail through lower liquefied soil by defining underlying soils as “bedrock.”

Output for all slope stability analyses can be found in Appendix C.

The yield acceleration was estimated as the applied horizontal acceleration corresponding to a factor of safety equal to unity. The yield acceleration, the height and period of the associated sliding mass, and the associated spectral acceleration developed from the Site Response Analysis (Section 8) were used as inputs into the Bray and Travasarou [2007] regression model to estimate lateral displacements. Table 10-3 shows the stability scenarios evaluated including the computed yield accelerations and associated estimated lateral displacements. The shallow pseudo-static stability surfaces for each cross-section are shown on Figures 10-7 through 10-9. The deep pseudo-static stability surface for cross-section SB-SB’ is shown on Figure 10-10.

The estimated lateral spreading displacements for this method are presented in Table 10-3. These values range from approximately 37 to 55 in. for shallow failure surfaces and approximately 15 to 20 in. for deep failure surfaces.

The shallow failure surface results using Bray and Travasarou [2007] for sections SA-SA’ and SB-SB’ (45 in. and 37 in. respectively) are larger than those predicted using Zhang et al. [2004] because the critical surfaces analyzed were for shorter distances


from the free face (~100 ft vs. 150 ft) and are therefore not representative of deformations affecting the Site structures. The larger 55 in. displacement computed for section SC-SC’ is a result of the section geometry (Figure 10-9) where there is a long “plateau” north of the perimeter berm which is at a lower elevation than the rest of the Site. This lower elevation results in reduced confining soil thickness above the liquefied zone, and hence reduced strength and greater susceptibility to lateral spreading. We note however, that the project structures are at a greater distance from the free face at section SC-SC’ and hence would anticipate lower magnitudes of lateral displacement at the structures along the northern boundary.

Summary of Lateral Spreading Displacement Estimates

Geosyntec employed two methods to estimate potential lateral spreading displacement at the Site, namely the strain potential approach and the Newmark sliding block approach. Both approaches indicate that lateral spreading displacements may occur at the Site during a liquefaction event. Based on Geosyntec’s analyses, potential lateral spreading displacements affecting the Site structures may be in the range of approximately 15 in. to 38 in. toward the Huntington Beach Flood Control Channel. The lower bound (15 in.) of this range corresponds to the deeper surfaces from the Newmark sliding block approach, while the upper bound (38 in.) corresponds to the strain potential method analyses at the closest approach of the Site structures to the channel (150 ft). It is recommended that the Project’s Structural and Geotechnical Engineers collaborate on the design of the foundation system to accommodate the anticipated lateral spread displacements. Several geotechnical mitigation methods that can reduce the magnitude of lateral spread displacements to acceptable levels are described briefly in Section 10.3.

10.3 Liquefaction and Lateral Spread Mitigation Strategies

Based on results of the investigation and analysis, no changes to the project layout are recommended to mitigate effects of liquefaction-induced settlements or lateral spreading displacements on stability of the proposed Site structures. However, project design measures against liquefaction-induced settlements and lateral spreading displacements are necessary to enhance structure stability.

Numerous options for design against liquefaction and lateral spreading are available. Examples of design measures include ground improvement and/or deep foundation


systems. Ground improvement may include over-excavation and re-compaction, deep soil mixing, jet grouting, and in-situ soil densification. Examples of deep foundation systems may include stone columns, rammed aggregate piers, and piles. These measures target an increase in soil density to reduce liquefaction settlement potential beneath critical structures, and an increase in soil strength to reduce the magnitude of lateral spread potential.

Based on the foregoing, Geosyntec recommends the following design measure to mitigate the impacts of the anticipated settlements and lateral spread displacements

• Design Measure B: During the Project’s design phase, Geosyntec recommends that the Project’s Structural and Geotechnical Engineers collaborate on the design of a foundation system for the proposed structures that can accommodate (1) approximately 9 inches of liquefaction-induced settlement; and (2) lateral spread displacement of approximately 15 to 38 inches. Such foundation system may include both geotechnical ground improvement methods to reduce the anticipated settlements and displacements to acceptable levels, and structural design methods to allow Site structures to tolerate the estimated settlements and displacements.


11. TSUNAMI HAZARD ASSESSMENT

Geosyntec recognizes that the Site is located within a mapped tsunami inundation area identified by Cal EMA [2009]. Therefore, Geosyntec considers an assessment of potential tsunami impacts at the Site to be warranted.

Previously, the CCC Staff [July 13, 2012] offered recommended values (i.e., run-up height, projected sea level rise, etc.) for assessment of potential tsunami impacts at the Site. Geosyntec [2013] reviewed these CCC Staff [July 13, 2012] recommendations and the supporting documentation. As described in detail in Appendix D, Geosyntec does not consider these CCC Staff [July 13, 2012] recommendations to be adequately supported. Therefore, Geosyntec considers the additional evaluation documented below to be needed.

As discussed in the remainder of this section, Geosyntec performed a review of the relevant technical literature and an evaluation to assess the potential tsunami impacts at the Site. Figure 11-1 illustrates schematically the key concepts described throughout this section and in Appendix D.

11.1 Baseline Tsunami Water Level12

Following the methodology presented during the technical exchange meeting between CCC Staff and Geosyntec on February 27, 2013 [Geosyntec, 2013], Geosyntec used elevation data available in Google Earth (www.google.com) to evaluate the baseline tsunami water level elevation implied by the location of the Cal EMA [2009] tsunami inundation line. This methodology is illustrated in Figure 11-2. As illustrated in the figure:

• An overlay of the Cal EMA [2009] map was created in Google Earth;

• The approximate limit of the mapped tsunami inundation line was delineated;

• A representative elevation along the approximate limit of the mapped tsunami inundation line was selected; and

12 See discussion of CCC Staff [July 13, 2012] recommendations in Appendix D of this report.


• This representative elevation was assumed as the baseline tsunami water level elevation.

Using this methodology, a baseline tsunami water level elevation of approximately 10 ft msl was selected for the Site. As indicated in the Cal EMA [2009] map, this value is already “…adjusted to ‘Mean High Water’ sea-level conditions, representing a conservative sea level…”

11.2 Projected Sea Level Rise over Design Life13

According to a report by the NRC [2012], the elevation of the local sea level is projected to rise during the life of the proposed facilities. This projected sea level rise varies by location. Of the locations mentioned in the NRC [2012] report, Los Angeles, California, is the closest to the Site. Therefore, projected sea level rise at Los Angeles, California, is assumed to be representative of sea level rise at the Site.

The NRC [2012] projections of sea level rise in the Los Angeles, California, area are shown in Figure 11-3. Also shown in Figure 11-3 is the planned design life of the proposed facility. Based on discussions with Poseidon, Geosyntec assumed the planned design life of the proposed facility to be approximately 30 years. As a projected date for the end of construction was not available as of the Report date, Geosyntec assumed facility construction to be completed between the years 2015 and 2020, Based on the assumed design life (i.e., 30 years) and the assumed end-of-construction date (i.e., 2015 to 2020), Geosyntec assumed that the design life of the proposed facility would end in the year 2050.

As shown in Figure 11-3, the NRC [2012] upper bound value of projected sea level rise in the last year of the planned design life (i.e., in the year 2050) is approximately 2.0 ft. Therefore, a value of 2.0 ft of sea level rise over the design life is assumed for the Site.

Geosyntec considers this value to be conservative. The selected value of 2.0 ft represents one “extreme” in that corresponds to the last year (i.e., the highest projected sea level rise) of the design life of the proposed facility. Concurrently, the selected value represents another “extreme” in that it corresponds to the upper bound of the projections presented in the NRC [2012] report.

13 See discussion of CCC Staff [July 13, 2012] recommendations in Appendix D of this report.


11.3 Sea Level Rise-Adjusted Tsunami Water Level

Adding the sea level rise projection of 2.0 ft to the baseline tsunami water level elevation of approximately 10 ft msl gives a sea level rise-adjusted tsunami water level elevation of approximately 12 ft msl.

11.4 Tsunami Inundation Potential

Based on the proposed finished floor elevations shown in Figure 11-4, the proposed Site improvements are planned to have finished floor elevations ranging from 9.0 ft msl to 14.0 ft msl. This range of finished floor elevations suggests that portions of the Site may be inundated by a tsunami with water level elevation 12 ft msl.

As shown in Figure 11-4, the planned Site configuration includes approximately 14 ft-high berms, with top elevations of approximately 22.3 feet MSL, around portions of the proposed facilities. However, as the current design does not provide for continuous berms around the Site, portions of the Site could be inundated by up to 3 ft of water (i.e., 12 ft msl minus 9 ft msl) during the tsunami event considered here.

It is noted that this evaluation does not consider effects of nearby construction outside the Site. For example, located between the ocean and the Site, new facilities may be constructed at the nearby AES HBGS. These facilities might have an effect on Site tsunami impacts (e.g., structures might perform a function similar to the existing berms, possibly reducing tsunami water velocity in portions of the Site).

Additionally, where the existing 14 ft-high berms reduce the tsunami water velocity, structures within the berms might experience reduced levels of tsunami-induced loading relative to comparable structures outside the berms.

11.5 Tsunami Impact at the Site

Possible impacts of a tsunami inundating portions of the Site may be divided into two categories: (i) possible impacts to Site improvements; and (ii) possible impacts to public health and safety.

Possible tsunami impacts on proposed Site improvements related to inundation of up to 3 ft of water include seepage, soil erosion, and loading on proposed structures. The impact of seepage is anticipated to be small as water inundation will be temporary. The


soil erosion impact also is likely to be small much of the Site is anticipated to be covered with concrete or asphalt pavement. Tsunami-related loading on the proposed structures can be mitigated by structural design.

Site tsunami hazard is not anticipated to pose a significant risk to public health and safety. According to the project’s SEIR, there will be fewer than twenty personnel on site during desalination plant operation. It should be noted that the Orange County Grand Jury [County of Orange, 2008] report on Tsunami Hazards finds that the City of Huntington Beach has one of the most advanced tsunami early-warning systems in the County. The City of Huntington Beach [2011] is also developing a Local Hazard Mitigation Plan that addresses tsunami risk to public health and safety.

11.6 Tsunami Impact Mitigation Strategies

Tsunami-related hazards at the Site related to public health and safety and facility structural stability can be mitigated. In order to mitigate the potential tsunami impacts at the Site, the following design measures are recommended:

• Design Measure C: Geosyntec recommends the implementation of SEIR mitigation measure HWQ-3: Prior to issuance of grading permits, the applicant shall submit to the City for approval a plan outlining the specific planning measures to be taken to minimize or reduce risks to property and human safety from tsunami during operation. Planning measures could include but would not be limited to the following: (a) Provision of tsunami safety information to all facility personnel, in addition to posting signage on site; (b) identification of the method for transmission of tsunami watch and warnings to facility personnel and persons on the site in the event a watch or warning is issued; and (c) identification of an evacuation site for persons on site in the event of a tsunami warning;

• Design Measure D: Geosyntec recommends the development of a coordinated Emergency Response Plan with AES HBGS prior to the commencement of Project operations; and

• Design Measure E: Geosyntec recommends the incorporation of tsunami-resistant design features into the design of proposed structures that are sufficient to accommodate potential inundation of up to approximately 3 ft. of water. Guidance on tsunami-resistant design that can sufficiently accommodate these


inundation levels and provide for vertical evacuation, if necessary, is available in the Applied Technology Council report titled Guidelines for Design of Structures for Vertical Evacuation from Tsunamis [ATC, 2008].


12. RECOMMENDATIONS

Geosyntec’s recommendations developed based on the geotechnical Site investigation and engineering evaluations documented in this Report are as follows:

• The Site is potentially liquefiable and is therefore considered a Site Class F per CBC [2010] and requires a site-specific seismic response analysis to develop a design acceleration response spectrum. This analysis was performed and results are shown in Figure 8-2. Because the computed average acceleration response spectrum is lower than the code-based minimum spectrum, CBC [2010] requires that the 80% Site Class E spectrum (shown on Figure 8-2) should be used for design at this Site.

• No mitigation of fault rupture hazard at this site is required. The 200 – 300 ft thick terrace deposit mitigates this hazard, even in an extreme case of a Mw 7.1 event induced rupture directly beneath the site.

• Based on Geosyntec’s evaluation of the potential for soil liquefaction and lateral spread on the Site, design measures against soil liquefaction and its impacts are required. Examples of design measures that may be implemented at the Site include, but are not limited to, installation of stone columns, installation of pile foundations designed to resist loads induced by liquefaction and lateral spreading, or application of one or more other ground improvement methods.

• Portions of the Site may be inundated during the tsunami event considered here by water depths up to approximately 3 ft. Mitigation measures related to health and safety of on-Site personnel are recommended to be implemented. These measures may include, but are not necessarily limited to, implementation of SEIR Mitigation Measure HWQ-3 and development of a Site-specific emergency response plan. Additionally, it is recommended to implement design measures for tsunami-related impacts on proposed structures. These measures may include elements of tsunami-resistant structural design.


13. LIMITATIONS

The conclusions and recommendations presented in this report apply to the Site conditions as we have described them and are the result of engineering studies and our interpretations of the existing geotechnical conditions at the time of our field activities. If any variations or undesirable conditions are encountered during future phases of work, Geosyntec should be notified so that supplemental recommendations can be developed if needed. The recommendations are based on the proposed Huntington Beach Seawater Desalination Project as described herein. If any significant changes to the project are proposed, those changes should be reviewed and the recommendations updated as needed.

The conclusions and recommendations presented in this report were developed consistent with generally accepted professional consulting principles and practices. No other warranty, express or implied, is made. Geosyntec is responsible for the conclusions and recommendations presented in this report for the project as described herein. We are not responsible for conclusions and recommendations made by others based on the information in this report, unless we are provided with an opportunity to review and concur with those conclusions and/or recommendations in writing. We do not warrant the accuracy of information supplied by others (e.g. boring logs, cross-sections, etc.) or the use of segregated portions of this report.


14. REFERENCES

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ASCE [2006], “Minimum Design Loads for Buildings and Other Structures,” ASCE Standard ASCE/SEI 7-05, American Society of Civil Engineers, Reston, Virginia.

ATC [2008], “Guidelines for Design of Structures for Vertical Evacuation from Tsunamis,” Report FEMA P646, Applied Technology Council, Redwood City, California.

Barrows, A.G., [1974], “A Review of the Geology and Earthquake History of the Newport-Inglewood Structural Zone, Southern California,” Special Report 114, California Division of Mines and Geology, Sacramento, California.

Bray, J. and Travasarou, T. [2007]. “Simplified Procedure for Estimating Earthquake Induced Deviatoric Slope Displacements” ASCE Journal of Geotechnical and Geoenvironmental Engineering, Vol. 133(4), p. 381-392.

Bray, J. and Travasarou, T. [2009] “Pseudostatic Coefficient for Use in Simplified Seismic Slope Stability Evaluation,” Journal of Geotechnical and Geoenvironmental Engineering, September 2009, v. 135(9), p. 1336-1340.

Burger, H.R., [1992], “Exploration Geophysics of the Shallow Subsurface,” Prentice Hall, Englewood Cliffs, New Jersey, 489 p.

Cal EMA [2009], “Tsunami Inundation Map For Emergency Planning, State of California - County of Orange, Newport Beach Quadrangle,” California Emergency Management Agency, Los Alamitos, California.

CBC [2010], “California Code of Regulations, Title 24, Part 2, Vol. 2 of 2,” Building Code (Based on 2006 International Building Code), California Building Standard Commission, Sacramento, California.

CCC Staff [2012], “Notice of Incomplete Coastal Development Permit (CDP) Application #E-06-007 - Poseidon Resources proposed Huntington Beach Desalination Facility,” dated July 13, 2012, Letter, California Coastal Commission, San Francisco, California.


CDMG [1997], “Guidelines for Evaluating and Mitigating Seismic Hazards in California,” Special Publication 117, California Department of Conservation, Division of Mines and Geology, Sacramento, California.

Chu, D.B., Stewart, J.P., Youd, T.L., and Chule, B.L. [2006] “Liquefaction-induced lateral spreading in near-fault regions during the 1999 Chi-Chi, Taiwan earthquake,” Journal of Geotechnical and Geoenvironmental Engineering, ASCE, Vol. 132, No. 12, pp. 1549-1565.

City of Huntington Beach [1996], “The City of Huntington Beach General Plan,” City of Huntington Beach, California.

City of Huntington Beach [2011], “Draft Hazard Mitigation Plan,” dated December 22, 2011, City of Huntington Beach, California.

County of Orange [2006] “County of Orange Emergency Operations Plans, Tsunami Annex,” County of Orange, California.

County of Orange [2008], “PARADISE LOST: If a Tsunami Strikes the Orange County Riviera...,” Grand Jury Report, County of Orange, California.

Eisner, R.K., Borrero, J.C., and Synolakis, C.E., [2001], “Inundation maps for the State of California,” In Proceedings of the International Tsunami Symposium 2001 (ITS 2001), NTHMP Review Session, R-4, Seattle, Washington, 7–10 August 2001, pp. 67-81.

Geosyntec [2012] “Geotechnical Hazards Assessment Report, Huntington Beach Seawater Desalination Project, Huntington Beach, California,” Project Number WG1665, dated December 2012, Technical Report, Geosyntec Consultants, Inc., San Francisco, California.

Geosyntec [2013] “Huntington Beach Desalination Project California Coastal Commission Briefing, Tsunami Hazard,” presented February 27, 2013, Technical Presentation to the California Coastal Commission Staff, San Francisco, California.

GLA [2002] “Preliminary Seismic Assessment.” Orange County Desalination Project, Huntington Beach, California, prepared for Poseidon Resources Corporation.

Houston, J.R., and Garcia, A.W. [1974], “Type 16 Flood Insurance Study,” WES Report H-74-3, United States Army Corps of Engineers.


Idriss, I.M. and Boulanger, R.W. [2008] “Soil Liquefaction During Earthquakes,” Earthquake Engineering Research Institute (EERI), Monograph 12.

IBC [2009], "International Building Code," International Code Council, Inc., Country Club Hills, Illinois.

IPCC [2007], “Climate Change 2007: The Physical Science Basis,” Contribution of Working Group I to the Fourth Assessment Report of the IPCC, S. Solomon, D. Qin, M. Manning, Z. Chen, M. Marquis, K.B. Averyt, M. Tignor and H.L. Miller, eds., Cambridge University Press, Cambridge, United Kingdom and New York, NY, USA, 996 pp.

Ishihara, K. [1986], "Evaluation of Soil Properties for Use in Earthquake Response Analysis," In: Geomechanical Modeling in Engineering Practice, R. Dungar and J.A. Studer, Eds., A.A. Balkema, Rotterdam, the Netherlands, 241 - 275.

Ishihara, K., and Yoshimine, M., [1992]. Evaluation of settlements in sand deposits following liquefaction during earthquakes, Soils and Foundations 32(1), 173–88.

LA DPW [2000], “Manual for Preparation of Geotechnical Reports,” Technical Guidance Document, County of Los Angeles, Department of Public Works, Alhambra, California.

Lazarte, C.A. [1996], “The Response of Earth Structures to Surface Fault Rupture,” Ph.D. Thesis, University of California at Berkeley, Berkeley, California.

Magorien, Scott [2002a], “Preliminary Review of Geotechnical Constraints and Geologic Hazards”, Appendix H to Draft Recirculated Environmental Impact Report for the Seawater Desalination Project

Magorien, Scott [2002b], “Preliminary Review of Geotechnical Constraints and Geologic Hazards, North and West Tank Options”, Appendix I to Draft Recirculated Environmental Impact Report for the Seawater Desalination Project

Magorien, Scott [2010], “Updated Preliminary Review of Geological Constraints and Geologic Hazards, Huntington Beach Desalination Report”, Appendix C to Draft Subsequent Environmental Impact Report for the Seawater Desalination Project


Matasovic N. [1993], “Seismic Response of Composite Horizontally-Layered Soil Deposits,” Ph.D. Dissertation, Civil Engineering Department, University of California, Los Angeles, 483 p. (www.GeoMotions.com).

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Matasovic, N. and Vucetic, M. [1995], “Seismic Response of Soil Deposits Composed of Fully-Saturated Clay and Sand,” Proc. 1st International Conference on Earthquake Geotechnical Engineering, Tokyo, Japan, Vol. 1, pp. 611-616.

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Robertson, P.K. and Cabal, K.L. [2010] “Guide to Cone Penetration Testing for Geotechnical Engineering,” Gregg Drilling and Testing Inc., 4th Edition, July 2012

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Salgado, R. [2008], “The Engineering of Foundations,” McGraw-Hill, New York, New York.

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Vucetic, M. and Dobry, R. [1991], “Effect of Soil Plasticity on Cyclic Response,” Journal of the Geotechnical Engineering, ASCE, Vol. 117, No. 1, pp. 89-107.

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TABLES

Total DepthCumulative Thickness of

Liquefied Soil

Cumulative Reconsolidation

Settlement1

(ft) (ft) (in)CPT-01 97.1 12.9 9CPT-02 96.1 8.7 6CPT-03 97.1 13.8 9CPT-042 50.0 3.8 3

CPT-052 50.0 2.3 21Reconsolidation Settlement estimated from Yoshimine et al. (2006)

Table 10-1. Liquefaction Induced Settlement

CPT

2CPT-04 and CPT-05 show less settlement because they were not advanced through all liquefiable materials

Total Depth

Cumulative Thickness of Liquefied Soil

under 50 ft bgs

Lateral Displacement Index (LDI)2

Lateral Displacement 150 ft from Free Face3

(ft) (ft) (in) (in)CPT-01 97.1 4 44 38CPT-02 96.1 4 43 37CPT-03 97.1 3 30 26CPT-041 50.0 4 39 33

CPT-051 50.0 2 33 281CPT-04 and CPT-05 driven to a maximum depth of 50 feet

3Free face height of the canal equal to 13 feet

CPT

Table 10-2. LDI-Based Lateral Spread Displacement

2Lateral Displacement Index (LDI) estimated from Zhang et al. (2004) considering only liquefiable soils within the first 50 feet

Coefficient of Yield Acceleration, ky

Lateral Displacement1

(g) (in)SA-SA' 1 2.4 -- --SA-SA' 2 1.0 0.03 45SA-SA' 3 1.0 0.11 15SB-SB' 1 3.1 -- --SB-SB' 2 1.0 0.04 37SB-SB' 3 1.0 0.11 15SC-SC' 1 2.5 -- --SC-SC' 2 1.0 0.02 55SC-SC' 3 1.0 0.09 20

1Estimated from Bray and Travasarou (2007)2Case 1 = Static; Case 2 = Shallow Psuedo-Static; Case 3 = Deep Psuedo-Static

Section CaseMinimum Factor of

Safety

Table 10-3. NEWMARK-BASED LATERAL SPREAD DISPLACEMENT

FIGURES

0 120'

N

B-1

CPT-3

LEGEND

PREVIOUS CONE PENETROMETER

TEST LOCATION [GLA, 2002]


TEST LOCATION [N&M, 2012]

PREVIOUS MUD ROTARY BOREHOLE

LOCATION [GLA, 2002]

PREVIOUS BOREHOLE LOCATION

[N&M, 2012]

CONE PENETROMETER TEST

LOCATION

EXISTING TOPOGRAPHY

CROSS SECTION ANALYZED

CPT-01

N:\C

AC

AD

D\S

\S

EA

WA

TE

R D

ES

AL

IN

AT

IO

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FILE NO.

FIGURE NO.

DATE:

PROJECT NO.2-1

1665F001-2

SITE LAYOUT

HUNTINGTON BEACH SEAWATER DESALINATION PROJECT

HUNTINGTON BEACH, CALIFORNIA

DECEMBER 2012

WG1665-02

A A'

CPT-1

B-2

WG1665-02\Figure 4-1 - Geologic Map.docx

GEOLOGIC MAP

HUNTINGTON BEACH SEAWATER DESALINATION PROJECT HUNTINGTON BEACH, CALIFORNIA

DATE: NOVEMBER 2012 FILE NO. PROJECT WG1665-02 FIGURE 4-1

Project Location

Not to Scale

Source: Rogers [1965]

WG1665-02\Figure 4-2 - Geologic Cross Section.docx

GEOLOGIC CROSS SECTION



Project Location

Not to Scale

Source: OCWD [2004]

WG1665-02\Figure 4-3 Idealized Soil Profile.docx

IDEALIZED SOIL PROFILE


DATE: NOVEMBER 2012 FILE NO. PROJECT NO. WG1665-02 FIGURE NO. 4-3

FILL γt = 125 pcf; Su = 1000 psf

SANDY CLAY γt = 110 pcf Su = 500 psf

SILTY SAND γt = 130 pcf; Φ = 35 deg

SILTY SAND γt = 120 pcf Φ = 35 deg

SILTY SAND: γt = 130 pcf; Φ = 35 deg


DEPTH (ft bgs)

ELEVATION (ft -MSL)

SANDY CLAY: γt = 110 pcf; Su = 1200 psf

10 ft

0 ft

-60 ft

-40 ft

-70 ft

0 ft

-2 ft

GWT 8 ft (Elev. +2 ft-MSL)

NOT TO SCALE

-63 ft

-73 ft

10 ft

70 ft

50 ft

80 ft

12 ft

73 ft

83 ft SANDY CLAY: γt = 110 pcf; Su = 1200 psf

WG1665-02\Figure 6-1 - Deaggregated Seismic Hazard.docx

DEAGGREGATED SEISMIC HAZARD FOR THE SITE



WG1665-02\Figure 6-2 - Acceleration Response Spectrum.docx

ACCELERATION RESPONSE SPECTRUM - BEDROCK



WG1665\Figure 6-3 - Deterministic SHA for Mw 7.5.docx

RESULTS OF DETERMINISTIC SEISMIC HAZARD ANALYSIS FOR MW 7.5


DATE: MARCH 2013 FILE NO. PROJECT WG1665 FIGURE 6-3

0.01

0.10

1.00

0.00 0.01 0.10 1.00

Peak

Hor

izont

al A

ccel

erat

ion

in B

edro

ck (g

)

Site-to Source Distance (km)

A & S (2008)

B & A (2008)

C & B (2008)

C & Y (2008)

Idriss (2008)

Median (5 Models)

Mw = 7.5Median PHGA = 0.58 g

WG1665-02\Figure 7-1 - Design Ground Motions.docx

DESIGN GROUND MOTIONS



WG1665-02\Figure 8-1 - Modulus Reduction and Damping.docx

MODULUS REDUCTION AND DAMPING CURVES



WG1665-02\Figure 8-2 - Acceleration Response Spectra.docx

ACCELERATION RESPONSE SPECTRA – SURFACE RESPONSE



WG1665-02\Figure 8-3 - Calculated Peak Shear Stress.docx

CALCULATED PEAK SHEAR STRESS PROFILE



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FILE NO.

FIGURE NO.

DATE:

PROJECT NO.9-1

1665F001-2

CROSS SECTION A-A'



OCTOBER 2012

WG1665-02

0 100'

WG1665-02\Figure 9-2 - FE Model Geometry.docx

FAULT RUPTURE IMPACT SIMULATION - FINITE ELEMENT MODEL


DATE: NOVEMBER 2012 FILE NO. AFW PROJECT NO. WG1665-02 FIGURE NO. 9-2

Elev. = 10 ft

Elev. = -200 ft

Horizontal Boundary Zero Horizontal and

Vertical Displacement Vertical Boundary

Zero Horizontal Displacement Horizontal Boundary Geostatic Step: Zero Horizontal and Vertical Displacement Load/Displacement Step: -0.67 ft Horizontal and -0.67 ft Vertical Displacement

Vertical Boundary Geostatic Step: Zero Horizontal Displacement Load/Displacement Step: -0.67 ft Horizontal and -0.67 ft Vertical Displacement

Elev. = 14 ft

+x

+y

Not to Scale

Water Table

Elev. = 2.6 ft

Bedrock 940 ft 1,560 ft

Soil-Concrete Interface

Elev. = -200 ft

Elev. = 10 ft

Subgrade Soil

Water Table

Subgrade Soil

Concrete Pad

Concrete Pad

WG1665-02\Figure 9-3 - Main Trace Displacement.docx

MAIN FAULT TRACE DISPLACEMENT MAGNITUDE


DATE: NOVEMBER 2012 FILE NO. AFW PROJECT WG1665-02 FIGURE 9-3

Average Surface Displacement

approximately 3.8 ft

Magnitude approximately 7.1

Source: Wells & Coppersmith [1994]

Average Subsurface Displacement approximately equal to Average Surface Displacement

WG1665-02\Figure 9-4 - Secondary Fault Displacement.docx

SECONDARY FAULT DISPLACEMENT MAGNITUDE


DATE: NOVEMBER 2012 FILE NO. AFW PROJECT WG1665-02 FIGURE 9-4

Note: After Lazarte [1996].

0

10

20

30

40

50

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70

80

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100

0 1 2 3 4 5 6 7 8 9 10

Dis

plac

emen

t div

ided

by

Dis

plac

emen

t on

Mai

n Fa

ult (

%)

Distance from Main Fault (mi)

Observations for Strike-Slip Secondary Faults Observations for Strike-Slip Branch FaultsObservations for Normal Secondary Faults Observations for Normal Branch FaultsObservations for Reverse Secondary Faults Observations for Reverse Branch FaultsObservations for Right-Normal Secondary Faults 25% of Main Fault Displacement

Secondary Fault Displacement approximately 25% of Main Fault Displacement

WG1665-02\Figure 9-5 - FE Model Results.docx

FAULT RUPTURE IMPACT SIMULATION - RESULTS


DATE: DECEMBER 2012 FILE NO. AFW PROJECT NO. WG1665-02 FIGURE NO. 9-5

1/100

1/1,000

1/10,000150 200 250 300 350 400 450 500 550 600 650 700 750 800 850

Calc

ulat

ed F

ound

atio

nDi

ffere

ntia

l Set

tlem

ent

x-Coordinate (ft)

Calculated Values ULS (Frame Cracking) SLS (Wall Cracking) SLS and ULS Unlikely

-0.

7

-0.

6

-0.

5

-0.

4

-0.3

-0.2

-0.1

0

Plot of Calculated Foundation Angular Distortion

Deformed Mesh (Displacement

Exaggerated 20x)

Approximate Limits of Proposed Structure

Approximate Limits of Proposed Structure

Contours of Vertical Displacement (ft)

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FILE NO.

FIGURE NO.

DATE:

PROJECT NO.10-1

1665F002-1

SITE LAYOUT WITH STABILITY CROSS-SECTIONS



DECEMBER 2012

WG1665-02

0 200'

N

B-2

CPT-3

LEGEND








[N&M, 2012]


LOCATION

EXISTING TOPOGRAPHY

STABILITY CROSS SECTIONS

ANALYZED

CPT-01

SA SA'

CPT-1

B-1

WG1665-02\Figure 10-2 Evaluation of Liquefaction Triggering.docx

EVALUATION OF LIQUEFACTION TRIGGERING



0.0 0.5 1.0 1.5 2.0Factor of Safety Against Liquefaction




-100

-80

-60

-40

-20

0

200.0 0.5 1.0 1.5 2.0

Elev

atio

n (ft

-MSL

)

Factor of Safety Against Liquefaction

CPT-01 CPT-02 CPT-03 CPT-04 CPT-05

LIQUEFACTION TRIGGERING BASED ON IDRISS AND BOULANGER (2008)

FS = 1.2

CPT TERMINATED AT 50 FT BGS

(~ELEV. 40 FT-MSL)


(~ELEV. 40 FT-MSL)

FS = 1.2 FS = 1.2 FS = 1.2 FS = 1.2

WG1665-02\Figure 10-3 Post-Liquefaction Idealized Soil Profile.docx

POST-LIQUEFACTION IDEALIZED SOIL PROFILE



LIQUEFIABLE SOILS ABOVE 50 FT BGS CONSIDERED FOR EVALUATION OF

LATERAL SPREAD AND PSUEDO-STATIC SLOPE

STABILITY

LIQUEFIABLE SOILS BELOW 50 FT BGS INCLUDED IN

SETTLEMENT EVALUATION

FILL γt = 125 pcf; Su = 1000 psf

SANDY CLAY γt = 110 pcf Su = 500 psf

SILTY SAND γt = 130 pcf; Φ = 35 deg


SILTY SAND: γt = 130 pcf; Φ = 35 deg


DEPTH (ft bgs)

ELEVATION (ft -MSL)

SANDY CLAY: γt = 110 pcf; Su = 1200 psf

10 ft

0 ft

-60 ft

-40 ft

-70 ft

0 ft

-2 ft

UPPER LIQUEFIED ZONE

(4-ft Thick) γt = 130 pcf

Su liq/σ’v = 0.05

LOWER LIQUEFIED ZONE γt = 130 pcf; Su liq/σ’v = 0.10

GWT 8 ft (Elev. +2 ft-MSL)

NOT TO SCALE

-6 ft

-35 ft

-63 ft

-73 ft

10 ft

70 ft

50 ft

80 ft

12 ft

16 ft

45 ft

73 ft

83 ft SANDY CLAY: γt = 110 pcf; Su = 1200 psf

WG1665-02\Figure 10-4 Evaluation of Post-Liquefaction Strength.docx

EVALUATION OF POST-LIQUEFACTION STRENGTH



-100

-80

-60

-40

-20

0

200.00 0.05 0.10 0.15 0.20

Elev

atio

n (ft

-MSL

)

Normalized Liquefied Strength, su(LIQ)/σ'vo

-100

-80

-60

-40

-20

0

200.00 0.05 0.10 0.15 0.20

Elev

atio

n (ft

-MSL

)


-100

-80

-60

-40

-20

0

200.00 0.05 0.10 0.15 0.20

Elev

atio

n (ft

-MSL

)


-100

-80

-60

-40

-20

0

200.00 0.05 0.10 0.15 0.20

Elev

atio

n (ft

-MSL

)


-100

-80

-60

-40

-20

0

200.00 0.05 0.10 0.15 0.20

Elev

atio

n (ft

-MSL

)



RESIDUAL STRENGTH BASED ON IDRISS AND BOULANGER (2008)

Su(LIQ)/σ’v = 0.05

Su(LIQ)/σ’v = 0.10

Su(LIQ)/σ’v = 0.05

Su(LIQ)/σ’v = 0.10

Su(LIQ)/σ’v = 0.05

Su(LIQ)/σ’v = 0.10

Su(LIQ)/σ’v = 0.05 Su(LIQ)/σ’v = 0.05


(~ELEV. 40 FT-MSL)


(~ELEV. 40 FT-MSL)

WG1665-02\Figure 10-5 Liquefaction Induced Settlement.docx

LIQUEFACTION INDUCED SETTLEMENT



-100

-80

-60

-40

-20

0

200 2 4 6 8 10

Elev

atio

n (ft

-MSL

)

Vertical Settlements (in), Sv (B) and Sv1D (R) 0 2 4 6 8 10




Vertical Settlements (in), Sv (B) and Sv1D (R)


LIQUEFACTION INDUCED SETTLEMENTS BASED ON ISHIHARA AND YOSHIMINE (1992)

Sv (BLUE) = INCREMENTAL SETLLEMENT Sv1D (RED) = CUMULATIVE SETTLEMENT


(~ELEV. 40 FT-MSL)


(ELEV. 40 FT-MSL)

CUMULATIVE SETTLEMENT

INCREMENTAL SETTLEMENTS

WG1665-02\Figure 10-6 LDI-Based Lateral Spread Displacement.docx

LDI-BASED LATERAL SPREAD DISPLACEMENT



-100

-80

-60

-40

-20

0

200 10 20 30 40 50

Elev

atio

n (ft

-MSL

)

Lateral Displacement Index, LDI (in)

SEE TEXT FOR ESTIMATIONS OF LATERAL SPREAD

0 10 20 30 40 50Lateral Displacement Index, LDI (in)





LATERAL DISPLACEMENT INDEX (LDI) BASED ON ZHANG ET AL. (2004)

A MAXIMUM OF 50 FT BGS (~ELEV. 40 FT-MSL) USED FOR LATERAL SPREAD ANALYSIS





WG1665-02\Figure 10-7 Section SA-SA' - Shallow Psuedo-Static Stability Surface.docx

SECTION SA-SA’ – SHALLOW PSUEDO-STATIC STABILITY SURFACE



0.98

Poseidon Slope Stabil i tySpencerP:\PRJ2003Geo\Poseidon\Slope Stabi l ity\Section A_upper l iq_ky_no wal l .gszHorz Seismic Load: 0.03

Name: Fi ll - Sandy Clay Model: Undrained (Phi=0) Unit Weight: 125 pcfCohesion: 1000 psf

Name: Plastic Clay Model: Undrained (Phi=0) Unit Weight: 110 pcfCohesion: 500 psf

Name: Upper Liquefied Sand Model: S=f(overburden) Unit Weight: 130 pcfTau/Sigma Ratio: 0.05 Minimum Strength: 0

Name: Bedrock Model: Bedrock (Impenetrable)





Distance0 50 100 150 200 250 300 350 400 450 500 550 600 650

Ele

vatio

n

-200

-175

-150

-125

-100

-75

-50

-25

0

25

50

ky = 0.03

WG1665-02\Figure 10-8 Section SB-SB' - Shallow Psuedo-Static Stability Surface.docx

SECTION SB-SB’ - SHALLOW PSUEDO-STATIC STABILITY SURFACE



0.98

Poseidon Slope Stabil i tySpencerP:\PRJ2003Geo\Poseidon\Slope Stabi l ity\Section B_upper l iq_ky.gszHorz Seismic Load: 0.04

Name: Fi l l - Sandy Clay Model : Undrained (Phi=0) Unit Weight: 125 pcfCohesion: 1000 psf

Name: Plastic Clay Model : Undrained (Phi=0) Unit Weight: 110 pcfCohesion: 500 psf


Name: Bedrock Model : Bedrock (Impenetrable)

Distance0 50 100 150 200 250 300 350 400 450 500 550 600 650

Ele

vatio

n

-200

-175

-150

-125

-100

-75

-50

-25

0

25

50

ky = 0.04

WG1665-02\Figure 10-9 Section SC-SC' Shallow Psuedo-Static Stability Surface.docx

SECTION SC-SC’ – SHALLOW PSUEDO-STATIC STABILITY SURFACE



1.01

Poseidon Slope StabilitySpencerP:\PRJ2003Geo\Poseidon\Slope Stability\Section C_upper liq_ky.gszHorz Seismic Load: 0.02

Name: Fill - Sandy Clay Model: Undrained (Phi=0) Unit Weight: 125 pcfCohesion: 1000 psf






Distance (x 1000)0.0 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 0.9

Elev

atio

n

-200

-175

-150

-125

-100

-75

-50

-25

0

25

50

ky = 0.02

WG1665-02\Figure 10-10 Section SB-SB' Deep Psuedo-Static Stability Surface.docx

SECTION SB-SB’ – DEEP PSUEDO-STATIC STABILITY SURFACE



1.00

Poseidon Slope StabilitySpencerP:\PRJ2003Geo\Poseidon\Slope Stability\Section B_upper liq+lseam_ky.gszHorz Seismic Load: 0.11

Name: Fill - Sandy Clay Model: Undrained (Phi=0) Unit Weight: 125 pcfCohesion: 1000 psf



Name: Poorly Graded Sand Model: Mohr-Coulomb Unit Weight: 130 pcfCohesion: 0 psfPhi: 35 °

Name: Lower Liquefied Sand Model: S=f(overburden) Unit Weight: 120 pcfTau/Sigma Ratio: 0.1 Minimum Strength: 0


Distance0 50 100 150 200 250 300 350 400 450 500 550 600 650 700

Elev

atio

n

-200

-175

-150

-125

-100

-75

-50

-25

0

25

50

ky = 0.11

WG1665\Figure 11-1 - Tsunami Key Terms.docx

TSUNAMI HAZARD ANALYSIS KEY TERMS


DATE: MARCH 2013 FILE NO. AFW PROJECT NO. WG1665 FIGURE NO. 11-1

Not to scale. For illustration purposes only.

Elevation Datum

Baseline Tsunami Water Level Elevation

Inundation Depth

Finished Floor Elevation

Site

Run-Up Height plus Tide Level

Projected Sea Level Rise

Sea Level Rise-Adjusted Tsunami Water Level

WG1665\Figure 11-2 - Baseline Tsunami Elevation.docx

BASELINE TSUNAMI WATER LEVEL ELEVATION


DATE: MARCH 2013 FILE NO. AFW PROJECT NO. WG1665 FIGURE NO. 11-2

CalEMA Tsunami Inundation Map Source: Cal EMA [2009]

CalEMA Tsunami Inundation Map

Overlay in Google Earth

Approximate Limits of Tsunami Inundation Zone

Site Approximate Location

Elevation 10 ft msl

WG1665\Figure 11-3 - Sea Level Rise.docx

PROJECTED SEA LEVEL RISE OVER DESIGN LIFE


DATE: MARCH 2013 FILE NO. AFW PROJECT WG1665 FIGURE 11-3

Note: Projections shown based on NRC [2012].

2.0

0.0

1.0

2.0

3.0

4.0

5.0

6.0

2000 2010 2020 2030 2040 2050 2060 2070 2080 2090 2100

Proj

ecte

d Se

a Le

vel R

ise

near

Los

Ang

eles

(ft

)

Year

"Mean""Mean ± One Standard Deviation""Range" based on IPCC [2007]Value Selected for Evaluation

Design Life of Proposed Facility

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FILE NO.

FIGURE NO.

DATE:

PROJECT NO.11-4

1665F001-2



MARCH 2013

WG1665

0 120'

N

B-2

CPT-3

LEGEND








[N&M, 2012]


LOCATION

EXISTING TOPOGRAPHY

DESIGN ELEVATION

CROSS SECTION ANALYZED

APPROXIMATE LOCATION OF

EXISTING BERM TO REMAIN IN

PLACE

APPROXIMATE LOCATION OF

EXISTING BERM TO BE REMOVED

CPT-01

A A'

9.00

CPT-1

B-1

Geotechnical Hazards Assessment Report

Documents

Transcript of Geotechnical Hazards Assessment Report