Finall Reportt 5 27-05-13 6

Chapter 1INTRODUCTION

1.1 General

In many countries, it is a common practice to construct RC frame building with open

ground storey (i.e. unlike other stories, no or scanty infill walls are provided in the

ground storey) in order to generate parking space, gardening space, and other utility

spaces for various purposes. Providing parking spaces in multi-storey buildings is an

essential requirement. Architect finds an easy solution by keeping ground storey open.

Also, the local municipal/building bylaw at many places supports/directs the same for

solving the parking problem. This is leading to a large number of open ground storey

building construction.

Generally, in open ground storey buildings, unreinforced brick masonry infills are

present in all floors except the ground story. This leads to severe stiffness and strength

irregularity and even sometimes leads to torsion irregularity. Buildings with these

irregularities has consistently shown poor performance during past earthquakes like

1999 Turkey, 1999 Taiwan and 2001 Bhuj, 2003 Algeria earthquakes and many

others. Normally, infill walls are considered as non-structural member; however,

practically it provides significant stiffness under lateral load. If special provisions

have not been followed in design, absence of infill at ground storey will lead to

formation of soft ground storey. Under lateral loading, lack of infill stiffness will lead

to larger inter-storey drift concentrated to ground storey leading to an early formation

of plastic hinges, further impending collapse of structure.

As per IS1893 (Part I): 2002 storey is considered as soft if its lateral stiffness is less

than 70% of that in the storey immediately above or less than 80% of the combine

stiffness of the three stories above. Also, an extreme soft storey is one in which the

lateral stiffness is less than 60% of that in the storey above or less than 70% of the

average stiffness of the three stories above.

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1.2 Literature Review on Behavior of RC Frame Building during Past

Earthquake

Past earthquakes have revealed the weaknesses of RC framed buildings and indicated

the points where improvements are required. In the following sub-sections some

damages to RC frame buildings have been highlighted including the damage and

behavior of open ground storey buildings and infill walls. Past earthquakes has

revealed the weaknesses of RC framed buildings and indicated the points where

improvements are required.

1.2.1 Seismic behavior of improperly designed, detailed and constructed RC frame

buildings

Due to ease of material handling and economy, the construction of RC frame

buildings in urban areas of the world is increasing day by day from last 4 decade.

Earlier, the main emphasis of design was for gravity loads, but the poor performance

of these gravity load designed buildings during earthquake has indicated the

importance of consideration of earthquake forces. In India, seismic codes are present

from last half decade, but due to lack of stringent guidelines and penalty, its

implementation in real construction is still not fully achieved. Gravity load designed

buildings suffers various types of damages during earthquake, following are few

commonly observed damages.

1) Brittle failure of RC column (Figure 1.1)

2) Short column failure (Figure 1.2)

3) Soft storey formation (Figure 1.3)

4) Weak column strong beam design (Figure 1.4)

5) Lap-splice failure (Figure 1.5 (a) and (b))

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Figure 1.2. Typical Earthquake Damage-Short Column Failure (1999 Athens Earthquake)

Figure 1.1 Brittle failure Photo from: Housner & Jennings, Earthquake Design Criteria, EERI, USA

Figure 1.3. Earthquake Damage- Dislodged Column due to Soft ground Floor effect (1999 Athens Earthquake)

Figure 1.4.The three-storyprimary school in Gedikbulak village after collapse (photos: Erdil) Turkey Earthquake

Figure1.5 (a).Lap splice joint fail Figure 1.5 (b).Closer view of lap splice failure of Figure 5 (a)

Takashi K. et al. (2000) reported that during Turkey earthquake of 17 August 1999, in

Kocaeli city which was also the epicenter of the earthquake, about 35839 residential

and 5478 workplaces RC buildings were heavily damaged / collapsed. Moderate

damage were observed in 41100 residences and 5861 work places, and slightly

damage in 45111 residences and 6122 workplaces all being constructed by reinforced

concrete. In another city Sakarya which is about 50 km from Kocaeli, number of

heavily damaged / collapsed RC buildings were 29844, moderately damaged were

22170 and slightly damaged were 26772.

After 12 January 2010 Haiti earthquake a survey 107 RC frame buildings conducted

by Eberhard et al. (2010) in Port-au-Prince downtown indicated that 28% had

collapsed and another 33% were damaged enough to require repairs. Another survey

conducted in Leogane city by the same author with 52 buildings, found that 62% had

collapsed and another 31% required repairs.

Issue of cumulative residual damage in buildings arises when the building is subjected

to high intensity multiple seismic loading during its designed life. The effect of

cumulative loading has been observed by Kam and Pampanin (2011) during two

consecutive earthquakes, first one on 4 September 2010 and second on 22nd February

2011. It has been observed that at Christchurch, the damage due to first earthquake is

confined to a moderate level, however, during the second earthquake many buildings

were severely damaged and about 135 buildings collapsed.

In India, the shortcomings of planning, design, detailing, and construction was

revealed during Bhuj earthquake. Bhuj earthquake of January 2001 was the first

earthquake in India which has affected urban area. Several thousand poorly designed

and constructed buildings were damaged and many of them collapsed. It was also

observed that the collapses of buildings were not limited only to epicentral region but

seen at Ahmadabad too, which is about 250 km away from the epicenter. In

Ahmadabad 75 RC frame buildings collapsed and several thousand were damaged in

and around the city, clearly demonstrating the seismic vulnerability of poorly

designed building (Jaiswal, et al.2003). In recent past, similar behavior of RC frame

buildings was also observed during Sikkim earthquake and Andaman earthquake.

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1.2.2 Seismic behavior of open ground storey buildings

Open ground storey buildings are those in which ground storey is kept open (free

from infill walls) to provide parking and other utility spaces in the building. In India

this is one of the most prevailing multi-storey construction practices. If open ground

storey buildings are improperly designed, it will have severe stiffness and strength

deficiencies. It has been observed during past earthquakes (1999 Turkey, 1999

Taiwan and 2001 Bhuj earthquake, 2003 Algeria earthquakes ) the damage in these

type of buildings are confined to the ground storey.

Earlier, this type of construction practices was advocated by various building by-laws.

The Taiwan government enacted a law to encourage contractor to construct buildings

with open ground storey (Tung and George, 2003). It was also instructed in the law

that ground storey height should be kept at least 5 m and in return the owners were

awarded with extra floor area.

Some common features buildings designed in Taiwan after the aforementioned law

was enacted are:

1. Generally ground storey was kept open with double height i.e. with a net

height approximately 7.6 meters (Figure 1.6 a).

2. Upper stories were having dense partition walls.

3. Due to presence of staircases and elevator shaft on the edge of the building

plan, torsional irregularity was also present.

4. Peculiar shapes of building plan were conceived by architect to provide good

outside view from every part of the building, natural lighting and ventilation.

5. To maximize parking space very few columns were designed into these

buildings at the basement. The primary load resisting system is reinforced

concrete moment resisting frame on a mat foundation.

The consequences of this law was revealed during 1999 Chi-Chi earthquake, where a

huge damage was observed in open ground storey buildings (Figure 1.6 b).

In India open ground construction is quite prevalent from last 25 years and its adverse

effect was observed during Bhuj earthquake. In Ahmadabad alone about 25,000, five-

storey buildings and about 1,500 eleven-storey buildings were damaged and about

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100 open ground storey buildings collapsed (Murthy 2005). Figure 1.7a and 1.7b

shows similar damage of open ground storey buildings during 2003 Boumerdes

earthquake and 2001 Bhuj earthquake, respectively. Further, in India, a large number

of similar open ground storey buildings exist in the various towns and cities situated

in moderate to severe seismic zones namely.

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Figure 1.6a. Double storey heighted open ground storey building in Taiwan (photo from World Housing Encyclopaedia Report)

Figure 1.6b. Collapsed Double storey heighted open ground storey building in Taiwan during Chi Chi earthquake 1999 (photo from World Housing Encyclopedia Report)

Figure 1.7a. Building collapse due to soft-story mechanism in the 2003 Boumerdes Earthquake (WHE Report 103, Algeria)

Figure 1.7b. Damaged RC frame building during Bhuj earthquake with damage confined to open ground storey (photo from EQ tips)

1.2.3 Seismic behavior of URM infilled frame

Unreinforced masonry (URM) infill walls are generally considered as non-structural

element. However, it has been observed that the behavior of infilled frame

significantly vary in comparison to bare frame under lateral loading. Modelling of

“Frame-Wall interaction” has remained a difficult task due to various reasons, such as

opening in wall, gap between wall and frame, and variation of material strength along

with significant increase in computational effort. A simplest modelling technique is

based on equivalent strut model. Masonry infill walls generally acts as a compression

strut when subjected to lateral loading as shown in figure 1.8. This model, initially

proposed by Poliakov consists of assuming that the effect of the infill panels can be

represented by introducing diagonal bars under compression. The existence of infill

walls can change the structural behavior from flexural action into axial action. Typical

failure mode of Infill wall and frame is shown in Figure 1.9. Failure modes include

corner crushing, frame damage, shear slip of wall, toe crushing, diagonal tension, etc.

The Advantages (Tabeshpour, et al., 2011) in the conversion of flexural action to axial

action are:

1) Reduce contribution of frame in lateral resisting

2) Reducing the lateral deformations

The Disadvantages (Tabeshpour, et al.2011) of converting the flexural action to axial

action:

1) Increase of the axial load in the column and foundation,

2) Creation of the concentrated shears at top and bottom of the column,

3) Creation concentrated shears at beginning and end of the beam

4) Creation of huge shears on the foundation

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Figure 1.8 Behaviour of infill wall subjected to lateral load. (photo from Klingner 1976)

Figure 1.9 Typical failure mode of URM infill wall and frame. (photo from Klingner 1976)

Figure 1.10 Failure of masonry walls during Turkey earthquake, 1999: (a) out-of-plane failure, (b) in-plane failure and (c) combined in- and out-of-plane failures (photo from Klingner 1976)

Figure 1.10 (a), (b), and (c) shows poor performance of infills during Turkey

earthquake. In Indian design practice the effect of infill in design is ignored, as a

result in Sikkim 14 Feb 2006 earthquake most RC buildings at Gangtok suffered

damages in some form or the other. The most common damage observed was cracks

in masonry infills, and separation between RC frame and infill. Not only private

society but also important structure like Government buildings including legislative

assembly building, Tashiling Secretariat, State Legislators’ Hostel, Geological Survey

of India (GSI) building at Deorali, suffered varying degree of damages. Among

Government buildings, GSI building was the worst affected. Fortunately, the ground

shaking was quite moderate and no RC building collapsed. And also in eastern and

southern Sikkim in private and government building infill and RC frame damage was

observed. In Sichey there was a government school a part of recently constructed

three-storey was found to be damaged. One masonry infill wall in the first storey

tilted out of plane (Figure 1.11 a) along with cracks in several other infills. (Kaushik,

et al.2006)

Figure 1.11. Government secondary school building at Sichey suffered moderate damages(Sikkim earthquake):(a) out of plane tilting of masonry infill wall, (b) inadequate shear reinforcement in columns, and spalling of cover concrete in columns at several locations due to corrosion

1.3 National and International Code provision for Infill Walls

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1.3.1 Indian Standard IS 1893 (Part-1): 2002

In IS1893:2002 Criteria for earthquake resistance design of structures Part 1 general

provisions and buildings gives only two formulae for frame without infilled (Eqn. 1.1)

and for frame with infilled (Eqn. 1.2). On modelling aspect of infill code is silent.

Ta=0.075 h0.75For RC frame without infill (1.1)

Ta=0 .09 h

√d For RC Frame with infill

(1.2)

1.3.2 FEMA-356 / ASCE-41

Federal Emergency and Management Authority or American Society for civil

engineering-41 provides guideline to generate infilled walls models in RC frame for

analysis. The elastic in plane stiffness of a solid unreinforced masonry infill panel

prior to cracking shall be presented with an equivalent diagonal compression strut of a

width ‘a’ given by equation below. The equivalent strut shall have the same thickness

and modulus of elasticity as the infill panel it represents (Figure 1.12)

a= 0.175 (λ1 hcol ) -0.04 rinf (1.3)

where: λ1=[ Eme t inf sin 2 θ

4 E fe I col hinf ]14 (1.4)

hcol= Column height between centerlines of

beams, in.

hinf= Height of infill panel, in.

Efe= Expected modulus of elasticity of frame

material, ksi

Eme= Expected modulus of elasticity of infill

material, ksi

Icol= Moment of inertia of column, in4.

Linf= Length of infill panel, in.

rinf= Diagonal length of infill panel, in.

tinf= Thickness of infill panel and equivalent strut, in.

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Figure 1.12 Showing compression strut analogy and parameters (Photo from FEMA356)

1.3.3 Eurocode 8

In Eurocode 8 which is for Design of structures for earthquake resistance gives some

clause about infills walls as follows:

4.3.6 Additional measures for masonry infilled frames

4.3.6.3 Irregularities due to masonry infills

4.3.6.4 Damage limitation of infills

5.9 Local effects due to masonry or concrete infills

6.10.3 Moment resisting frames with infills

1.4 Literature review on retrofitting techniques

In recent years there has been a substantial advancement in research on repair and

seismic retrofitting of existing building as evident from increasing number of the

growing number of research papers published in this area.

Repair: The process to regain original strength of a damage or deteriorated structure is

called as Repair.

Seismic Retrofitting: The process to enhance original strength of a deficient or

damaged structure and enabling it to satisfactorily can perform its intended

performance in future seismic event is called retrofitting.

1.4.1 Objective of retrofitting

Objective of seismic retrofitting is only to improve seismic performance of building.

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1.4.2 Retrofitting strategies

Retrofitting strategy is the basic overall approach to enhance the probable seismic

performance of the building or to otherwise reduce the existing risk to an acceptable

level [ATC-40]. Retrofitting strategies can be categorized as: (1) Completion of load

path and removal of structural irregularity (2) Strengthening of structure (3)

Enhancing deformation capacity of structure and (4) Reducing earthquake demand.

A structure, which is deficient in original design, can be retrofitted by strengthening

of the structure. This strengthening can be achieved either by adding new lateral load

resisting members or by strengthening the existing members. Large number of

techniques based on conventional strengthening method, such as addition of new

members (shear walls, bracings), RC jacketing, steel jacketing, as well as, based on

advance material such as FRP have been developed. Strengthening is the most

suitable and commonly used method of retrofitting for URM infilled RC frames.

Deformation capacity of URM infilled RC frame buildings can also be increased to

some extent by improving the ductility of beams, columns and infills. However, this

approach has limited scope due to brittle behaviour of URM infills. Supplemental

energy dissipation does not have much utility in URM infilled RC frame buildings

due to low inter-story drifts. Base isolation is a promising approach useful for low rise

buildings. In the present review, focus will be on various seismic strengthening

techniques and effectiveness of stiffness and ductility enhancement. There are four

strategies of retrofitting which are as follows:

1) Stiffness increase

2) Strength increase

3) Ductility increase

4) Mass reduction

But we are mainly focusing on stiffness and strength increase of structure due to

retrofitting. While remaining two strategies, ductility and mass reduction is not

primary attention.

1.4.3 Retrofitting techniques

Retrofitting techniques are the specific methods used to implement the overall retrofit

strategy. Under a given retrofitting strategy a number of retrofitting techniques are

available. Typical load-displacement relationships for different strengthening

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techniques have been shown in Figure1.13. The figure shows that for monolithic wall,

strength of structure is very high but ductility is very low and for unstrengthened

frame column ductility is very high but strength is less. So it can be understood from

this Figure that both ductility and strength cannot be achieved at the same time.

Figure 1.13 Typical load-displacement relationships for different strengthening

Techniques [Rodriguez et. al. (1991)]

1.4.3.1 Addition of shear wall

Addition of shear wall (Figure 1.14 a) into an existing building is most common

approach of seismic retrofitting. It has been used with frame, since long time. It is an

effective method of increasing building strength and stiffness. When shear walls are

situated at proper positions in a building, they can form an efficient lateral-force

resisting system, while simultaneously fulfilling functional requirements. Addition of

shear wall improves buildings strength and stiffness, and also it is economically

feasible and readily compatible with most of existing concrete buildings. It also gives

good aesthetic view. Many buildings all over the world have been retrofitted using

shear walls. In Japan, from a period of 1933 to 1975 about 85% case of retrofitting

was executed using shear walls [Rodriguez et al. (1991)]. Similarly, in countries like

United States, India, Turkey, and other places use of shear wall for seismic retrofitting

of existing buildings is well accepted [Pincheira (1993), Holmes (2000), Moehle

(2000)].

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Various analytical models have been proposed in literature to simulate the behaviour

of RC shear wall. These include Equivalent beam model, three vertical line element

model, inelastic analytical macroscopic model, Macro-finite-element model.

In equivalent beam model the shear wall member is replaced at its centroidal axis by a

line element and connected by rigid link to the frame beams. The main limitation of

this model lies in the assumption that rotation occurs around points belonging to the

centroidal axis of the wall so it does not accounts for migration of the neutral axis of

the wall cross-section, rocking of the wall etc. In three vertical line element model, a

generic wall member is idealized as three vertical line element with infinitely rigid

beam at the top and bottom floor levels. Two outside truss elements represent the

axial stiffness of the boundary columns, while the central element was one-component

model with vertical, horizontal and rotational springs concentrated at the base. This

model is capable of describing flexural and shears deformation including the

migration of the neutral axis of the wall cross-section, rocking of wall etc., but the

deformation due to the fixed end rotation and web splitting-crushing mode of failure

is not accounted [Vulcano and Bertero 1987]. Inelastic analytical macroscopic model

uses eight inelastic axial springs connected by two rigid beams to account plastic

bending deformation of wall, and three shear springs which expresses the shear

behaviour of panel and two boundary columns [Fu, et al., 1992]. Macro-Finite-

Element model consists of a number of vertical elements. These vertical elements

consist of vertical and horizontal springs at the centre of each vertical element. The

axial stiffness of each vertical spring is represented by two parallel components

representing mechanical behaviour of the concrete and steel. Horizontal spring

represents the shear springs. The stiffness of each shear spring is determined by the

different state of vertical spring. In this model the axial springs first reach the

nonlinear state and then the shear spring, thus the effect of axial stiffness on shear

stiffness was neglected.

As reported by Mike Griffith in his JRC Scientific and Technical Reports a recent

experimental study by Altin et al (1992) tested fourteen 2-storey by 2-bay concrete

frames that were strengthened with concrete infill walls cast-in place. The

effectiveness of various degrees of inter-connection between the infill wall

reinforcement and the surrounding concrete frame were assessed and all results were

compared to the hysteretic behaviour of the bare concrete frame. It was observed that

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while the peak strength (reached at about 0.4% drift) was not sensitive to the degree

of frame wall connectivity, the hysteretic behaviour was best for the most integrally

connected infill walls (maximum displacements corresponding to drifts of

approximately 1 to 1.5%). Finally, in the recent European Conference on Earthquake

Engineering, 3 papers were presented on this topic. The first of these presented the

results of analyses on the seismic response of concrete frame school buildings in

Taiwan retrofitted with concrete infill walls [Sheu, et al., 1998]. The paper by Pop et

al (1998) presented the experimental results of tests on bonded anchors for use

between concrete infill walls, which were added to pre-existing concrete frames. It

was determined that an embedment length of 8 bar diameters into the concrete frame

was required to achieve optimal force transfer and interaction. The third paper

[Ozcebe, et al., 1998] presented the results of an experimental investigation into the

effectiveness of cast-in-place concrete walls as a seismic retrofit strategy for concrete

frame buildings damaged in the 1995 Dinar, Turkey earthquake. The tests showed that

the once the walls had been added the frame had little apparent effect on the strength

of the rehabilitated structure. However, lap splices in the columns prevented the infill

from achieving full effect. Steel plate jackets were recommended for the column

splice zones to solve this problem.

There are some adverse effects of addition of shear wall for retrofitting also we should

aware of this. If large number of shear wall added then it result in increase in mass of

the building and therefore increase in seismic forces also demand i.e., requirement of

strength increases. Shear walls can effect into architectural impact through the loss of

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(b)Figure 1.14 Shear wall (a) shear wall in building (b) Proper anchoring of

vertical reinforcement into foundation (Photo from Murthy CVR EQ tips)

(a)

windows. It also requires special foundation (Figure 1.14 b) work which highly

expensive as it produces large overturning forces at their base (ATC-40).

1.4.3.2 Addition of bracing

Additions of bracing to RC frames are another common method of seismic retrofitting

which increase stiffness, strength and ductility. But it provide less stiffness and

strength compare to shear walls. It can construct with less disruption in building with

very small loss of lights and have smaller effect of traffic patterns within building.

While strengthening a frame using bracing, the main unknown is how the new bracing

system will interact and behave when attached to an existing structure. Generally, it

has been found that bracing is an efficient technique and can be used in combination

with other techniques such as interior shear wall or column jacketing. Experimental

studies on strengthening of existing frames using bracing shows a significant

improvement in strength, stiffness and even ductility [Baboux, et al., 1990, Masri et

al., 1996]. It’s very difficult to attach braces with frame in seismic retrofitting. In

some studies, the bracings are directly fitted within concrete frame, whereas in some

cases steel frame is used to attach the bracing within concrete frame. Currently, there

is a lack of rational provision for analysing, designing and detailing these systems.

For simplicity, many times the direct summation of the strength and stiffness of

concrete frame and steel frame are done; this will often underestimate the total

strength. The total strength of the retrofit system should be determined as the

composite strength of the steel/concrete system [Moehle, 2000]. Ranges of bracing

system have been proposed in the literature for upgrading existing concrete frame.

This include concentric bracing (diagonal and X-bracing), Eccentric bracing, post-

tensioned bracing and buckling restrained bracing.

The main steps in retrofitting with steel bracing are outlined in the flowchart [Badoux,

M. and Jirsa, J.O. 1990] of (Figure 1.15) Retrofitting decision is based on an

inadequate evaluation of structure (step 1), If the structure is found inadequate (step 2)

and a retrofitting option is chosen (step 3), the engineer must determine the aim of the

retrofitting operation (step 4)—the structural and Architectural requirements. The

selection of the best retrofitting scheme (step 5) depends on the existing structure. If a

bracing scheme is selected, the first step of the design is the choice of the layout of the

system (step 6).Next, the members and connections are detailed (step 7). The

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foundations of the braced frame may require strengthening because greater foundation

forces are typically generated in the retrofitted frames under lateral loads. Finally, the

designer must try to anticipate particular construction problems that are likely to come

up when working on an existing structure (step 9). Allowance should be made for

fitting to clearances and possible in situ modification.

Figure 1.15 Flowchart for Retrofitting with Bracing System

1.4.3.2.1 Concentric Bracing

This type of bracing (Figure 1.16 a) has been widely used in strengthening of existing

frames [Nateghi; 1995]. Concentric bracings essentially consist of one or more

diagonal members. Experimental tests on strengthening existing frames using

concentric bracing showed its adequacy for lateral load resistance. Test on one

diagonal brace shows 2.5 times increase in shear strength, whereas X-bracing showed

4 times increase [Maheri, 1997]. In the same experiment, behaviour of X-braced

frame under the loading showed that the tension braces carries a large portion of load

and the failure of bracing starts with failure of tension braces followed by buckling

failure of compression braces. Other tests also showed that, 60% to 70% of the total

load applied was carried by braces before buckling [Bush et al., 1991a, Bush et al.,

1991b] and in the same experiment it has been observed that The maximum lateral

load applied to the strengthened frame was 2.24 times the predicted design capacity

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and also The initial stiffness of the braced frame (precracked) was 1.5 times that of

the uncracked original frame.

1.4.3.2.2 Eccentric Bracing

An eccentric brace (Figure 1.16 b) differs from a conventional brace in a way that, the

centerlines of eccentric braces do not intersect at the center of the beam column joint,

but rather are offset horizontally from the joint. Eccentrically braced frames (EBFs)

are now gaining acceptance in seismic strengthening of RC frames [Ghobarah et al.,

2001, Perera et al., 2004]. EBFs have been recognized as efficient technique for

enhancing seismic resistance because in addition to strength and stiffness it also

provides ductility. One major advantage of eccentric bracing is protection from

buckling under the higher loads generated by a major earthquake. One more benefit of

using this type of bracing system is the considerable flexibility regarding the

placement of braces. In this type of bracing system, the axial forces induced in the

braces are transferred either to a column or to another brace through shear and

bending in a segment of the beam. The critical beam segment is called an “active

link” or simply link. Thus in this type of system yielding and inelastic energy

absorption occurs at the links created by brace. Stiffness of EBFs ranges between

stiffness of moment resisting frames to concentric braced frame.

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Figure 1.16 Bracing(a) Concentric bracing Lorant G.

<http://www.wbdg.org/resources/seismic_design.php>

Figure 1.16 Bracing(b) Eccentric bracing <http://web.iku.edu.tr/courses/insaat/ce007/>

Therefore these types of bracing system are a hybrid deriving its stiffness from truss

action and its ductility by inelastic deformation of the link [Popov, et al., 1987]. Based

on behaviour of links, EBFS are categorized as eccentric shear braces and eccentric

flexural brace. If ductility of the link beam relies primarily on shear deformation, it is

usually referred to as an eccentric shear brace (shear capacity of a short link is usually

less than its flexural capacity). If the deformation of the link beam is dominated by

flexure, it is described as an eccentric flexural brace. From experimental investigation

conducted by Malley (1984), it was found that, active link which yield primarily in

shear are more effective energy dissipater than which yield primarily in bending.

1.4.3.2.3 Post-tensioned steel bracing

The use of post-tensioned steel braces in seismic rehabilitation is relatively new

technique that can be applied efficiently to existing low and middle-rise frame

buildings. This type of bracing consists of high-slenderness steel strands which can

take only tension and are subjected to a prestressing force. Different levels of

prestress, based on building property, soil condition and desired building behaviour

can be applied to post-tensioned braces [Tena-Colunga, 1996] Teran-Glimore, et al.,

1996]. An initial prestress of 75% of their yield strength was applied during

strengthening of a low rise building in US [Pincheria, 1993] to determine its

adequacy. Such a level of initial brace prestress will cause braces to yield and allow

energy dissipation at relatively small lateral. Behaviours of the post-tensioned braces

was idealized using a bi-linear model that considered yielding in tension and elastic

buckling in.

1.4.3.2.4 Buckling restrained bracings

Even though conventional concentric bracing systems are efficient, but they suffer

buckling due to high slenderness ratio. The problem of buckling has lead to

development of buckling restrained braces (BRBs) or unbounded braces (UBs). BRBs

or UBs basically consists of three components, a steel core member, buckling

restraining part and the unbonding material as shown in Figure 1.17(a) & (b).

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(a) (b)

Figure 1.17 (a) Schematic of BRBs or UBs, (b) Typical types of BRBs [Tsai, et al., 2004]

In these types of braces, a core steel cross-shapes or flat bar member is encased into a

steel tube and confined by infill concrete. The steel core member is designed to resist

the axial force with full tension or compression yield capacity without the local or

global flexural buckling failure. When the brace is subjected to compression, the

unbonding material placed between the core member and the infill reduces the

friction. Figure 3(b) illustrates the typical cross sections of BRBs.

1.4.3.3 Jacketing

Jacketing adds both strength and stiffness to structure. There are various types of

jacketing generally observed. Such as column jacketing, beam jacketing, infill

jacketing, column-beam joint jacketing, etc.

1.4.3.3.1 Column jacketing

In 1970’s earthquake many of the structural failures due to inadequate shear strength

and/or improper spacing in confinement in concrete columns. So, to increase column

cross section column strengthening procedures i.e., jacketing of column (Figure 1.18)

is used. The main problem with this approach is that it often unacceptably increases

the dimension of the column, rendering the retrofit impractical. The use of thin carbon

fibre composite sheets avoids this problem and has consequently gained acceptance

over the past 10 years. Nevertheless, concrete jacketing of concrete columns has been

shown to be very effective in improving strength and ductility and converting strong-

beam weak-column buildings into buildings with a strong-column weak-beam

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mechanism (Choudhuri, et al., 1992; Rodriguez and Park, 1994; Bracci, et al., 1997;

Bush, et al., 1990).

1.4.3.3.2 Beam column joint jacketing

As we saw previously in earthquake damages survey one of main reason of RC frame

damages are beam column joint failure. And is due to inadequate joint reinforcement

and/or improper anchorage of longitudinal beam reinforcement or improper provision

of development length. As reported by Mike Griffith Ghobarah, et al., (1996a,b)

presented this type of retrofitting method of jacketing to RC beam-column joints with

corrugated steel sheeting. Column beam joint jacketing was shown to not suffer from

the outward swelling problems of steel jacketing with flat steel sheets as reported by

Priestley (1994a). The corrugations act to stiffen the jacket in the needed “hoop”, or

tie, direction and so maintain good confinement of the concrete. Four concrete test

specimens were tested, one which represented existing structures, one represented

current seismic design standards and two rehabilitated connections. Results indicated

good cyclic performance at high load levels and significant increases in shear capacity

(V kN kN u = 400 → 450 ) and displacement ductility ( = 2→ 4 Δ μ ) (As reported by

Griffith M.).

And also infill jacketing is also use for retrofitting. In which shotcreting,

prefabrication of reinforced concrete panels attached with dowels through masonry

walls, steel plates apply on masonry walls.

21

Figure 1.18 Jacketing of column (Photo from Famer group

<http://www.famergroup.com/earthquake.html>)

http://www.famergroup.com/earthquake.html

1.4.3.4 Friction dampers

Retrofitting the structure with friction dampers provide the strength control required

to obtain the optimum structural response. Adding friction dampers increase stiffness

of the frame until a certain shear level is reached, at which the dampers can be set to

slip. For earthquake loading a appropriate slip level can be selected to give optimum

response. The energy dissipated by the friction dampers reduces the energy demand

on the structure and damps out the structural response. (Pall A.S. and Pall R. 1991)

The Friction brake is widely used to extract kinetic energy from a moving body as it is

the most effective mean to dissipate energy. More than centuries, mechanical

engineers had successfully used this concept and also currently using this concept to

control motion of machinery and automobiles. This principle inspired to develop

friction dampers. Patented Pall friction dampers are available for tension cross

bracing, single diagonal bracing, cladding connections and friction base isolators.

Before shifted to site of building it is well tested to ensure proper slip load. It is very

inexpensive i.e., economical in cost and easy to handle. Its construction is also easy

and easy in install. It’s made up of specially treated series of steel plates to develop

friction as shown in Figure 1.19. They are clamped together with high strength steel

bolts and these plates are allowed to move on each other with a pre define load.

Figure 1.19 Pall Friction Damper (Photo from Golafshani. A. A and Gholizad.A 2009)

It is an effective way to control seismic response of structure and non-structural

damage. It does not impact the foundation design, features that make them

22

particularly attractive for upgrading existing buildings. They also seem to offer

savings in the initial cost of new structures and retrofitting of the existing ones

compared to conventional solutions. However, it is very difficult to maintain its

properties for long time intervals. It also has other disadvantages like they are

effective only for flexible structure, they encumber the design procedure and make it

more expensive and the selection of the appropriate slip load is a critical issue in the

performance of a structure with friction dampers.

These dampers have successfully gone through rigorous testing on shake table in

Canada and US. As reported (Filistrault 1986) in 1985 a 3 storey frame equipped with

friction-dampers was tested on a shake table. And it was found that frame using

friction dampers gives result much superior to that of moment- resisting frame and

moment resisting braced frame. An earthquake record with peak acceleration of 0.9g

did not cause any damage to friction- damped braced frame. While other two cause

permanent deformation. Also in 1987, a 9 storey three bay frame, equipped with

friction dampers, was tested on a shake table at Earthquake Engineering Research

Centre of the university of California at Berkeley. All members of the friction-

damped frame remained elastic for 0.84g acceleration which is maximum capacity of

shake table, while the moment- resisting frame would have yield at about 0.3g

acceleration (Aiken, et al., 1988).

1.5 Objective of the study1) To compare period & base shear of different frame like bare frame, infill

frame, and open ground storey frame by IS code & SAP model.

2) Performance of evaluation of these building by non-linear static procedure.

3) Comparison of performance enhancement of these building with different

retrofitting techniques.

4) Identification of most suitable retrofitting techniques.

1.6 Methodology

1) Generic plan of RC frame building selected.

2) Building will be model for different height i.e.,G+3, G+7, G+15

3) Selection of suitable modeling techniques in SAP

4) Modeling, design and comparison of base shear and period of vibration of

these building.

23

5) Performance of evaluation of these building by non-linear static procedure.

6) Selection of retrofitting techniques & corresponding modeling techniques in

SAP.

7) Base on result will identify most suitable retrofitting techniques.

1.7 Scope of the work

Work will be limited to one type of limited plan and three different heights.

24

Chapter 2

SELECTION OF GENERIC PLAN

2.1 Introduction

In this study, a generic plan of a typical college building has been selected. The plan

in one direction has large number of bays i.e., nine and in other direction it has three

bays. This type of plan has been selected to simulate different behaviour in

longitudinal and transverse direction. In longitudinal direction there are nine bays

which indicate sufficient redundancy. Where as in transverse direction there are only

three bays which represent comparatively lesser redundancy. In addition to this the

effect of infill walls will be prominent in longitudinal direction than in transverse

direction.

Figure 2.1 Showing plan and elevation of G+3 model

Figure 2.2 Showing Plan of G+3 bare frame model in SAP2000

25

Figure 2.3 Showing elevation of G+3 bare frame model in SAP2000

To calculate time period and acceleration by IS 1893:2002 take a generic plan of plan

dimension 41.4m X 14.7m shown in figure 2.1& 2.2 and in elevation (Figure 2.1 &

2.3 ) taking three building of G+3, G+7and G+ 15.

After design of G+3 building dimension of column, beam, slab, infill, and also load

on frame sections are

Section properties:

Column 1 (depth, width) =0.60 ×0.6 in meter

Column 2 (depth, width) =0.40 ×0.6 in meter

Beam 1 = 0.23x0.35 in meter



Slab (thickness) = 0.14 in meter

Outer infill wall (thickness) = 0.23 m

Inner infill wall (thickness) = 0.115 m

Load

Live load Roof floor =3.5 KN/m

Live load intermediate floors= 3.5 KN/m

2.2 Comparisons study

2.2.1 Time period Comparisons

In the present study, time period has been calculated as per IS1893:2002 using codal

provision method. It can be observed from Table 2.1 that while considering infill wall

26

effect time period reduces drastically as compared to the model without infill wall

effect.

Table 2.1 Comparison of time period with and without infilled wall model.

Time period (with infill wall)in Sec.

Time period (without infill wall) in Sec.

Storey X-Direction Y-Direction X and Y-Direction

G+3 IS-1893 0.233 0.373 0.597

G+7 IS-1893 0.440 0.711 0.968

G+15 IS-1893 0.865 1.387 1.598

2.2.2 Modal mass participation factors

It can be observed from Table 2.2 that the modal mass participation factors are well

distributed in bare frame and bare frame with infill load only. When model as bare

frame in SAP2000 mode shape are observed as 1st step in X direction 1st mode and 2nd

step in Y direction of a 1st mode. But when we model infill as a strut in SAP2000 then

x-direction first mode observed in step no.3 and Y-direction first mode in step 1 and

second step number is torsion mode it means the stiffness of model is changed due to

addition of infill wall. In Y direction stiffness is less compare to X direction as there

are only three bays in Y direction while 9 bays in X direction (cell filled with dark

colour).

Table 2.2 Comparison of Modal mass Participation factor by RSA in SAP 2000

Types of Building

Direction Mode No. 1 2 3 Steps no.

Bare frameX Modal Load

participation factor

0.83228 0.0919 0.02403 1'4'7'

Y 0.81587 0.10453 0.03005 2'5'9'

Bare frame with infill load only

X Modal Load participatio

n factor

0.84611 0.06364 0.02152 2'5'9'

Y 0.84029 0.09692 0.02432 1'4'7'

27

42

22

32

52

12

1

Full infilledX Modal Load

participation factor

0.8828 0.07065 0.01303 3'7'11'

Y 0.84107 0.09706 0.02433 1'4'6'

Open Ground storey

X Modal Load participatio

n factor

0.92287 0.02662 0.0042 3'6'10'

Y 0.84133 0.09694 0.02422 1'4'7'

2.2.3 Shear force and bending moment

Figure 2.4 Showing X-Z view of G+3 building in SAP 2000

In the present study, in Figure 2.4 different members of a building which have been

made bold are selected for calculating bending moment and shear force. Then these

moment and force with different models like bare frame, bare frame with infilled

load, full infilled and open ground storey building have been compared. From result it

has been concluded that if full infilled wall model take then both bending moment and

shear force increase (cell dark in table 2.3, 2.4, 2.5) compare to other frame model

i.e., bare frame and full infilled wall. And when we take member no.1 (table 2.3)

which is column of ground storey, shear force and bending moment are increase

compare to member 2 (table 2.4) which is column of first storey and increment is 2.14

times in shear force and 1.53 times in bending moment and also column no. 1 bending

moment and shear force increases compare to column element no. 4 (table 2.6). In

case of beam no. 3 compare to beam no. 5 bending moment and shear force increases

(table 2.5 & 2.7).

Table 2.3 Comparison of Shear force and Bending moment element No.1 shown in Figure 2.4 by RSA in SAP 2000

28

Type of Building

Location 1

CombinationsShear Force

in kNCombinations

Bending moment

kN m

Bare frame

Left side Corner ground storey

column of 1st frame

1.5(DL+EQx) 41 1.5(DL+EQx) 83

Bare frame with infill load

only1.5(DL+EQx) 77 1.5(DL+EQx) 196

Full infilled 1.5(DL+EQx) 56 1.5(DL+EQx) 112

Open Ground storey

1.5(DL+EQx) 176 1.5(DL+EQx) 321


Type of Building

Location 2

CombinationsShear Force

in kNCombinations

Bending moment

kN.m

Bare frame

Left side Corner 1st

storey column of 1st frame

1.5(DL+EQx) 30 1.5(DL+EQx) 61




Open Ground storey

1.5(DL+EQx) 56 1.5(DL+EQx) 211


Type of BuildingLocation

3Combinations

Shear Force in kN

CombinationsBending moment

kN.m

Bare frame

Left side Corner 1st

storey slab Beam of 1st frame

1.5(DL+EQx) 59 1.5(DL+EQx) 110




Open Ground storey

1.5(DL+EQX) 129 1.5(DL+EQX) 180

29


Type of Building

Location CombinationsShear Force

in kNCombinations

Bending moment

kN.m

Bare frame

Left side Corner 3rd

storey column of 1st frame

1.5(DL+EQx) 15 1.5(DL+EQx) 38



Full infilled 1.5(DL+EQX) 20 1.5(DL+EQX) 60

Open Ground storey

1.5(DL+EQX) 20 1.5(DL+EQX) 57

Type of Building

Location CombinationsShear

Force in KN

CombinationsBending moment KN.M.

Bare frame

Left side Corner 3rd

storey slab beam of 1st

frame

1.2(DL+LL+EQx) 32 1.2(DL+LL+EQx) 41


only1.2(DL+LL+EQx) 39 1.5(DL+EQx) 55

Full infilled 1.5(DL+LL) 33 1.5(DL+LL) 30

Open Ground storey

1.2(DL+LL+EQx) 28 1.2(DL+LL+EQx) 27


Chapter 3

30

MODELLING AND ANALYSIS

3.1 Introduction

Performance based seismic engineering is the modern approach to earthquake

resistance design. Rather than being based on prescriptive mostly empirical code

formulation performance based design is an attempt to predict building with

predictable seismic performance. Therefore, performance objective such as life safety,

collapse prevention and immediate occupancy are used to define the state of building

following a design earthquake. This chapter provides what is nonlinear analysis and

pushover analysis with actual nonlinear analysis result and conclusion.

3.2 Non-linear static procedure

To model the complex behaviour of reinforced concrete structure analytically in its

non-linear zone is difficult. This has led engineers in the past to rely heavily on

empirical formulas which were derived from numerous experiments for the design of

reinforced concrete structures. For structural design and assessment of reinforced

concrete members, the non-linear analysis has become an important tool. The method

can be used to study the behaviour of reinforced concrete structures including force

redistribution. This analysis of the nonlinear response of RC structures to be carried

out in a routine fashion. It helps in the investigation of the behaviour of the structure

under different loading conditions, its load deflection behaviour and the cracks

pattern. Simplified nonlinear analysis procedures using pushover analysis methods,

such as capacity spectrum method requires determination of three primary elements:

capacity, demand (displacement) and performance.

Capacity: Capacity is representation of the structures ability to resist the seismic

demand. The overall capacity of a structure depends on the strength and deformation

capacities of the individual components of the structure. In order to determine

capacities beyond the elastic limits, some form of nonlinear analysis, such as

pushover procedure is required. This procedure uses a series of sequential elastic

analysis, superimposed to approximate a force-displacements capacity diagram of the

overall structure. The mathematical model of the structures is modified to account for

reduced resistance of yielding components. A lateral force distribution is again

31

applied until additional components yield. This process is continued until the structure

becomes unstable or until a predetermined limit is reached. The pushover capacity

curve as shown in Figure 3.1, approximates the structures behave after exceeding their

elastic limit.

Figure 3.1 Capacity curve

Demand (Displacement): Demand is representation of the earthquake ground motion.

During an earthquake Ground motion produce complex horizontal displacement

patterns in structures that may vary with time. Tracking this motion at every time

steps to determine structural design requirements is judged impractical. Traditional

linear analysis methods use lateral forces to represent a design condition. For

nonlinear methods it is easier and more direct to use a set of lateral displacement

demand is an estimate of the maximum expected response of the building during the

ground motion. Figure 3.2 shows elastic response spectrum also called demand curve.

Figure 3.2.Demand curve

Performance: Once a capacity curve and demand curve is defined, a performance

check can be done. In other words, the structure must have a capacity to resist

32

Roof displacement

Bas

e sh

ear

Capacity curve

Elastic response spectrum spectrum

earthquake demand such that the performance of the structure is compatible with the

objective of the design. A performance check verifies that structural and non-

structural components are not damaged beyond the acceptable limits of the

performance objective for the forces and displacements implied by the displacement

demand. Figure 3.3 shows intersection of demand curve and capacity curve which is

called as performance point.

Figure 3.3 Performance point

3.2.1 Pushover analysis

ATC 40 and FEMA 273, FEMA 356 and FEMA 440 have described the Push Over

analysis procedure, modelling of different components and acceptable limits. In

general, this method develops a damage curve (capacity curve) for a given direction.

To get the desired performance level from this curve, two methods, namely, Capacity

Spectrum Method and Displacement Coefficient Method have been introduced in

FEMA 440. This analysis procedure considers only first mode shape of the equivalent

single degree of freedom system and predefined vertical distribution of the load along

height in one direction at a time. These are the limitations of this method. Still it is

very efficient analysis procedure because it gives full insight of the nonlinear behavior

of the structure.

Pushover procedure is used to determine capacity curve. It determines capacity

beyond elastic limit. It is basically a step by step plastic analysis for which the lateral

loads of constant relative magnitude are applied to a given structure and progressively

increased until a target displacement is reached, while gravity loads are kept constant.

33

Performance point

Pushover analysis consists of a series of sequential elastic analysis, superimposed to

approximate a force- displacement curve of the overall structure. A two or three-

dimensional model which includes bilinear or trilinear load-deformation diagrams of

all lateral force resisting elements is first created and gravity loads are applied

initially. A predefined lateral load pattern, which is distributed along the building

height, is then applied. The lateral forces are increased until some members yield. The

structural model is modified to account for the reduced stiffness of yielded members

and lateral forces are again increased until additional members yield. The process is

continued until a control displacement at the top of building reaches a certain level of

deformation or structure becomes unstable. The roof displacement is plotted with base

shear to get the global capacity curve. Figure 3.4 shows flow chart of pushover

analysis by Capacity spectrum method (ATC-40)

Figure 3.4 Flow chart of capacity spectrum method (ATC-40)

Pushover analysis can be performed as force-controlled or displacement controlled. In

force-control pushover procedure, full load combination is applied as specified, i.e.,

34

force-controlled procedure should be used when the load is known (such as gravity

loading). Also, in force controlled pushover procedure some numerical problems that

affect the accuracy of results occur since target displacement may be associated with a

very small positive or even a negative lateral stiffness because of the development of

mechanisms and P-delta effects.

Generally, pushover analysis is performed as displacement-controlled proposed to

overcome these problems. In displacement controlled procedure, specified drifts are

sought (as in seismic loading) where the magnitude of applied load is not known in

advance. The magnitude of load combination is increased or decreased as necessary

until the control displacement reaches a specified value. Generally, roof displacement

at the centre of mass of structure is choosing as the control displacement.

The practical difficulties associated with the non-linear direct numerical integration of

the equations of motion leads to the use of non-linear static pushover analysis of

structures. Pushover analysis is getting popular due to its simplicity. High frequency

modes and nonlinear effects may play an important role in stiff and irregular

structures. The contribution of higher modes in pushover analysis is not fully

developed.

3.2.2 Modelling and analysis procedure used in the present study

In present study, analysis performed using SAP2000 nonlinear version 14.2.4. A three

dimensional model of the structure has been define as a frame element. Define plastic

hinges at both ends of beams and column. The nonlinear properties of plastic hinges

are to be given as input to SAP2000.

3.2.2.1 Nonlinear modelling of Masonary infilled

Large number of research has been carried out in the past on analytical modeling of

masonry infills, Based on these studies, it was observed that masonry infills can be

conveniently modeled as compressive diagonal struts. In the present study, the elastic

in-plane stiffness of a solid unreinforced masonry infill panel prior to cracking shall

be represented with an equivalent diagonal compression strut of width. The equivalent

strut shall have the same thickness and modulus of elasticity as the infill panel it

represents shown in figure 1.12. It was reported that specimens with strong frames

35

and strong panels exhibited both better lateral load resistance and energy dissipation

capacity performance than those with weak frames and weak panels. The width of

compressive struts ‘a’ was considered as given in FEMA 356. The thickness of struts

was taken as the actual thickness of walls (Figure 1.12).

Determination of equivalent strut

Figure 3.6 Showing sectional view in YZ plane and infill model as strut in SAP2000

Strut width is depend on various factors such as span of bay, height of infill, thickness

of infill, quality of masonary, Size of infill, adjacent column width and depth etc., In

figure 3.6 beam size change in 1st and 2nd bay from 3rd floor to 2nd floor that’s why

strut width change S-J, S-K, S-M, etc., but in 3rd bay, beam size is similar in all floor

that’s why strut width is same and denoted by ‘S-L’. Some infills struts widths are

shown in below table 3.1 by FEMA 356.

Table 3.1 Calculated equivalent strut width for different infill

Strut Name

Equivalent strut width in m.

S-J 0.569

S-K 0.505

S-L 0.938

S-M 0.567

S-N 0.502

36

The analytical modeling of infilled frames is a complex issue because these structures

exhibit highly nonlinear inelastic behaviour resulting from the interaction of the

masonry infill panel and the surrounding frame. The modeling approaches for

masonry can be grouped into micro models and macro models. Micro models capture

the behavior of infill and its interaction with the frames in much detail, but these

models are computationally expensive and time consuming. On the other hand, macro

models try to capture the gross behavior of the infill, are approximate but

computationally efficient.

In this study, masonary act as a strut and model as a lumped plasticity model. If the

lateral force is applied on building then the force is developed in strut which acts like

compression. So that’s why axial hinge is provided at centre of strut. To calculate

axial hinge we require some nonlinear properties which are shown in figure 3.7 & 3.8

like yield force, displacement control parameter C, D, E and also acceptance criteria.

Figure 3.8 Nonlinear static procedure-simplified Force- Deflection Relations for Masonary

infill panels (photo from FEMA356)

37

Figure 3.7-b Manual hinge provision in SAP2000

Figure 3.7-a Manual hinge provision in SAP2000

To calculate yield force there are some formulae, given in FEMA 356 which are

F y= Vinecos θ

(N) (3.1)

V ine=A∋x f vie(N ) (3.2)

Where, Fy is maximum allowable yield force,

Vine is design shear force,

Ani is area of strut,

Fvie is expected shear strength of masonry infill,

And to calculate values of IO, LS & CP (figure3.9) there are some formulae given in FEMA356 and also from table 7-9 (figure 3.8) A strut=t inf X ESW (mm) (3.3)

d y=% X h inf X cosθ (3.4)

Dy= Py

Astrut x EmeLinf

(3.5)

PD=(d y−D y )

1000

(3.6)

d ls=(% for LS) X hinf X cosθ

(3.7)

Where,Astrut = Area of strut,

Eme = Expected elastic modulus of masonry in Compression,

Linf = Length of infill,

hinf = Height of infill,

θ = Angle between infill diagonal and horizontal axis,

tinf = thickness of infill,

ESW=Equivalent strut width.

38

Figure 3.9 Generalized Force-Deformation relations for masonry Elements or Components (Photo from FEMA356)

Table 3.2 Nonlinear properties of infill hinges

Infill name

Yield Force

Displacement control parameter

Acceptance Criteria

Fy (kN) B C D, E LS CP

S-J 261.80 0 0.083 0.083 0.063 0.083

S-K 231.66 0 0.089 0.089 0.069 0.089

S-L 421.04 0 0.026 0.026 0.016 0.026

S-M 260.84 0 0.083 0.083 0.063 0.083

S-N 230.58 0 0.089 0.089 0.069 0.089

Figure 3.10 Nonlinear properties of axial hinge to be filled in SAP

Table 3.2 shows the calculated nonlinear properties of some infills and figure 3.10

shows how to fill these properties in SAP2000.

39

3.2.2.2 Pushover analysis of G+3, G+7, G+15 stories

Pushover procedure is use d to determine capacity curve. It determines capacity

beyond elastic limit. It is basically a step by step plastic analysis for which the lateral

loads of constant relative magnitude are applied to a given structure (figure 3.11) and

progressively increased until a target displacement is reached, while gravity loads are

kept constant.

Figure 3.11 Static approximation used in the pushover analysis (photo from Mortezaei A.)

Figure 3.12 Load pattern in pushover analysis (photo from A. Mortezaei)

Pushover analysis consists of a series of sequential elastic analysis, superimposed to

approximate a force- displacement curve of the overall structure. A two or three-

dimensional model which includes bilinear or trilinear load-deformation diagrams of

all lateral force resisting elements is first created and gravity loads are applied

initially. A predefined lateral load pattern, which is distributed along the building

height, is then applied (figure 3.12). The lateral forces are increased until some

members yield. The structural model is modified to account for the reduced stiffness

of yielded members and lateral forces are again increased until additional members

yield. The process is continued until a control displacement at the top of building

reaches a certain level of deformation or structure becomes unstable. Figure 3.13

40

shows yielding sequences when lateral load applied. The roof displacement is plotted

with base shear to get the global capacity curve.

In present work, Pushover analysis of 4, 8 & 16 story building with bare frame, full

infill and open ground story model have been studied.

Figure 3.13 Yielding sequences through conventional pushover analysis (photo from A. Mortezaei)

Figure 3.15 shows Pushover curve of g+3 stories in which 3 curves of bare frame, full

infilled and open ground story model. Bare frame model shows a large deformation

and ductility. Due to formation of mechanism open ground story model gives very

less displacement and there is not deformation further and analysis is stops before

formation of collapse hinges.

In Y direction, G+3 stories pushover curve shows (Figure 3.16) large displacement

and high ductility in bare frame. When same frame modelled as a full infilled, curve

shows decrease in ultimate displacement and less ductility. Strength carrying capacity

of this full infilled model is increase because of high rigidity & stiffness due to infill

but ultimate displacement is decrease. In OGS Strength increase as compare to bare

frame but ductility and deformation decreases.

Conclusion: From pushover curve in Y direction we conclude that due to infill,

structure strength increase but ductility decrease as compare to bare frame.

Figure 3.17 shows pushover curve of G+7 stories. In bare fame X direction, as usual

curve shows high displacement due to high stiffness and also shows large ultimate

displacement. In case of full infill same as G+3, strength capacity of building

increases but ultimate displacement decrease and in OGS frame ductility is reduce

drastically and structure fails early as compare to G+3 building.

41

Conclusion: As we go towards high stories OGS construction is dangerous. So there is

requirement of some retrofitting.

In Y direction of G+7 stories building (figure 3.18), pushover curve of bare frame

shows large displacement and full infilled and OGS also shows displacement. There is

not so much difference in strength carrying capacity also.

Conclusion: As we see plan of building in figure 3.18, there is only four frame in Y

direction as compare to 8 stories so there is not that much difference in base shear and

displacement as we observe in X direction.

Figure 3.19 shows pushover curve of G+15 stories. In bare frame X direction, as usual

curve shows high deformation due to high stiffness and also shows large ultimate

displacement. In case of full infill same as G+3 & G+7, strength capacity of building

increases but ultimate displacement decrease and in OGS frame ductility is reduce

drastically and structure fails early.

For Y direction in figure 3.20 same as in X direction

Figure 3.14 Plan and elevation of G+3 RC frame Building

42

0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 0.40

2000

4000

6000

8000

10000

12000

14000

full in-fill

OGS

Bare frame

Displacement in m

Ba

se s

hea

r in

kN

B- IO- LS- CP- DBE- MCE-

Figure 3.15 Pushover curve of G+3 building in X - direction

0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 0.40

2000

4000

6000

8000

10000

Bare frame

Infill

OGS

Displacement in m

Ba

se s

hea

r in

kN


Figure 3.16 Pushover curve of G+3 building in Y- direction

0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 0.4 0.45 0.50

3000

6000

9000

12000

15000

18000

21000

Bare frame

Full infill

OGS

Displacement in m

Base

sh

ear

kN


Figure 3.17 Pushover curve of G+7 building in X- direction

43

0 0.1 0.2 0.3 0.4 0.5 0.60

2000

4000

6000

8000

10000

12000

Bare frame

Full in-fill

OGS

Displacement in m

Base

sh

ear

in k

N



0 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.80

2000

4000

6000

8000

10000

12000

14000

Full infill

OGS

Bare frame

Displacemnt in m

Bas

e sh

ear

in k

N


Figure 3.19 Pushover curve of G+15 building in X- direction

0 0.1 0.2 0.3 0.4 0.5 0.60

1000

2000

3000

4000

5000

6000

7000

Full infill

OGS

Bare frame

Displacementi in m

Ba

se s

hea

r in

kN



3.2.2.3 Bi-Linearization of curve

44

Figure 3.21 shows the bi-linearization method of curve, which is used in present study

to linearize pushover curve of all three type building. After linearization, calculate

ductility demand which is calculated by ratio of ultimate displacement (Δu) to the

yield displacement (Δy) and Fy and Fu which is yield force and ultimate force

respectively which are calculated by bi-linearization method and shown in table 3.3

for x- direction and in table 3.4 for y- direction.

Figure 3.21 Bi-linearization of curve (FEMA356)

3.1.2.4 Determination of ductility

Table 3.3 Ductility, yield force, ultimate force in bare frame, full infilled and open ground

story model of different building in X- direction.

G+3 StoryDuctility=

Δu/ΔyG+7 Story

Ductility

=Δu/ΔyG+15 Story

Ductility=

Δu/Δy

Bare frame

Fy = 4300kN

5.214Bare frame

Fy = 9800kN

4.273Bare frame

Fy=11997 kN

3.020

Δy = 70mm

Δy =110mm

Δy=245mm

Fu = 4486 kN

Fu=10000kN

Fu=11997 kN

Δu = 365mm

Δu =470mm

Δu =740mm

Full infilled

Fy = 6556 kN

1.000Full

infilled

Fy=16000kN

5.571Full

infilled

Fy=17100 kN

2.400

Δy = 20mm

Δy = 70mm

Δy =150mm

Fu = 6556 kN

Fu=18388kN

Fu=19118 kN

Δu = 20mm

Δu =390mm

Δu =360mm

45

Open ground story

Fy = 5400 kN

1.600Open

ground story

Fy=15200kN

2.000Open

ground story

Fy=14000kN

2.000

Δy = 30mm

Δy = 80mm

Δy =125mm

Fu = 5811 kN

Fu=16633kN

Fu=17346 kN

Δu = 48mm

Δu =160mm

Δu =250mm

Table 3.4 Ductility, yield force, ultimate force in bare frame, full infilled and open ground

story model of different building in Y- direction.

G+3 StoryDuctility=

Δu/ΔyStory G+7

Ductility=

Δu/ΔyStory G+15

Ductility= Δu/Δy

Bare frame

Fy =4100kN

6.182Bare frame

Fy=7950 kN

5.611Bare frame

Fy=2050kN

9.800

Δy =55mm

Δy =90mm

Δy=100mm

Fu=4927 kN

Fu=8708 kN

Fu=2182kN

Δu =340mm

Δu=505mm

Δu=980mm

Full infilled

Fy=6000 kN

3.000Full

infilled

Fy=8200 kN

6.800Full

infilled

Fy=6200 kN

4.211

Δy = 25mm

Δy = 50mm

Δy =95mm

Fu=7975 kN

Fu=10319k

N

Fu=7136kN

Δu =75mm

Δu=340mm

Δu=400mm

Open ground story

Fy=5800 kN

4.000Open

ground story

Fy=7900 kN

5.727Open

ground story

Fy=5400kN

2.167

Δy =30mm

Δy = 55mm

Δy = 90mm

Fu=7552 kN

Fu= 9963 kN

Fu=6200 kN

Δu =120mm

Δu =315mm

Δu =195mm

3.2.2.5 Determination of Performance point

For calculating performance point Displacement Modification Method (DMM) has

been used to obtain the performance point (FEMA 356). In the DMM, the target

displacement ɗt roof level can be as

46

ɗt=C0C1C2SaxT e2

4 X π 2 xg

Where, Co is modification factor to relate spectral displacement of an equivalent

SDOF system to the roof displacement of the building. C1 is a modification factor to

relate expected maximum inelastic displacement to displacement calculated for linear

elastic response. For period greater than 1.0s the value of C1 is taken as 1. C2 is

modification factor to represent the effect of pinched hysteresis shape, cyclic stiffness

degradation on maximum displacement response. For period greater than 0.7s the

factor C2 is 1. Sa is the spectral acceleration, at effective fundamental period Te and the

damping ration of the building in the direction under consideration; g is acceleration

due to gravity.

Figure 3.15 to 3.20 shows pushover curve with different performance level like B-

operational level in pink colour, IO- Immediate occupancy in blue colour, LS-Life

safety in faint blue colour, CP-Collapse prevention in green colour. Also shows

design base earthquake (DBE) by triangle sign in faint green colour and maximum

consider earthquake by square sign (MCE).

In figure 3.15, when DBE earthquake comes pink type hinge form in bare frame

means structure is in operational level and when MCE type earthquake comes then

structure is in immediate occupancy level in bare frame.

In figure 3.16, pushover in Y direction when DBE type earthquake come structure is

in operational level in bare frame but in full infill and OGS structure is in immediate

occupancy level and when MCE type earthquake comes full infill & OGS structure is

nearly collapse.

In figure 3.17, pushover in X direction of G+7 stories when DBE type earthquake

comes structure is in operational level in bare frame and in other type of model

structure is in IO level. When MCE type earthquake comes OGS structure is nearly in

collapse level and infill is in operational level. In Y Direction figure 3.18, when

consider MCE earthquake bare frame is in IO level, full infill & OGS is in life safety

level.

In figure 3.19, pushover curve in X direction of G+15 building when DBE type

earthquake comes structure is in operational level in bare frame and IO level in OGS

and full infill frame. When MCE type earthquake comes OGS is just near to collapse

level. In case of figure 3.20 in Y direction of a structure same.

47

3.2.2.6 Comparison of inter story drift ratio

Inter story drift ratio is very essential parameters for designing the non-structural

components of the building. In present work, It has been studied inter story drift ratio

in X & Y direction i.e., longer & short direction using pushover analysis. The

variation of inter story drift ratio at MCE level along the height of building is shown

in figure 3.28 to 3.33 and variation of inter story drift ratio at ultimate level is shown

in figure 3.22 to 3.27 . It has been observed that story drift demand is more in middle

stories in bare frame & full infilled frame model. The reason is sudden reduction of

cross section area of the frame elements along the height of the structures. When we

are considering infilled walls model drift demand has got substantially reduced. But in

case of open ground story model drift demand is more in ground story and as we go

above it decreases. Because there is no infill walls as result stiffness is reduces

drastically and it became soft story or weak story in both at MCE level and ultimate

displacement. For tall buildings, infills substantially affect the drift behaviour of the

buildings.

0.00 0.50 1.00 1.50 2.00 2.50 3.000

1

2

3

4

5

Bare frame

Full in-filled

OGS

Inter Story Drift Ratio (%)

Sto

ry le

vel

Figure 3.22 Inter Story Drift ratio in X-direction of 4 story at ultimate displacement.

48

0.00 0.50 1.00 1.50 2.00 2.50 3.00 3.500

1

2

3

4

5

Bare frame

Full in-filled

OGS

Inter Story Drift Ratio (%)

Sto

ry l

evel

Figure 3.23 Inter Story Drift ratio in Y-direction of 4 story at ultimate displacement.

0 0.5 1 1.5 2 2.5 30

1

2

3

4

5

6

7

8

9

OGS

Bare frame

Full infilled

Inter story Drift Ratio (%)

Sto

ry

lev

el


0 0.5 1 1.5 2 2.50

1

2

3

4

5

6

7

8

9

OGS

Bare frame

Full in-filled


Sto

ry le

vel


49

0.00 0.50 1.00 1.50 2.00 2.50 3.000

2

4

6

8

10

12

14

16

18

Bare frame

Full infill

OGS


Sto

ry

lev

el


0.00 0.50 1.00 1.50 2.00 2.50 3.00 3.500

2

4

6

8

10

12

14

16

18

Bare frame

Full infill

OGS


Sto

ry

lev

el


0.00 0.20 0.40 0.60 0.80 1.00 1.20 1.400

1

2

3

4

5

Bare frame

Full in-filled

OGS

Inter story drift Ratio (%)

Sto

ry l

evel

Figure 3.28 Inter Story Drift ratio in X-direction of 4 story at MCE.

50

0.00 0.20 0.40 0.60 0.80 1.00 1.200

1

2

3

4

5

Bare frame

Full in-filled

OGS


Sto

ry le

vel

Figure 3.29 Inter Story Drift ratio in Y-direction of 4 story at MCE.

0.00 0.10 0.20 0.30 0.40 0.50 0.60 0.700

1

2

3

4

5

6

7

8

9

OGS

Bare frame

Full in-filled


Stor

y le

vel


0 0.2 0.4 0.6 0.8 1 1.20

1

2

3

4

5

6

7

8

9

OGS

Bare frame

Full in-filled


Stor

y le

vel


51

0.00 0.50 1.00 1.50 2.00 2.50 3.000

2

4

6

8

10

12

14

16

18

Bare frame

Full infill

OGS


Sto

ry l

evel


0.00 0.50 1.00 1.50 2.00 2.50 3.000

2

4

6

8

10

12

14

16

18

Bare frame

Full in-fill

OGS


Stor

y le

vel


52

CHAPTER 4

SEISMIC RETROFITTING TECHNIQUES

4.1 Retrofitting techniques

As in previously study in non linear analysis the adverse effects of open ground story

like increase story drift ratio in ground story, Capacity curves gave small

displacement, hinges formations in ground story columns i.e., formation of

mechanisms and ground story columns fails before beams, etc. So these buldings are

requires special arrangements to increase the lateral stiffness & strength of the soft

story. In Indian code (IS 1893:2002 Part 1) there is provision for soft story

strengthening in clause 7.10.3 (a) which states that “The columns and beams of the

soft storey are to be designed for 2.5 times the storey shears and moments calculated

under seismic loads”(figure 4.1). While 7.10.3 (b) which states that “Placing shear

walls symmetrically in both directions of the building as far away from the centre of

the building as possible and it is to be designed for 1.5 times the lateral storey shear

force”.

53

Figure 4.1 OGS member designed for higher forces using code specified factors

Also other international code like Eurocode 8 CEN 2003 recommends increasing the

design forces for the soft first-story columns by 1.5 to 4.68 times & it depends upon

several factors. Israeli seismic code SII 1995 also recommends increasing design

forces of open first story and also for one adjacent story above by 2.1–3.0 times the

actual design forces, depending on the ductility level of the building. According to the

Bulgarian Seismic Code 1987, the seismic design forces for soft story in masonry-

infilled RC frames are required to be increased by 2 times the corresponding design

forces for a regularly infilled frame, and by three times the design seismic forces for a

regular bare frame. (Kaushik H.B.)

4.1.1 2.5 times increasing design forces of column & beam in soft story

The effectiveness of these strengthening schemes, in which the lateral strength of the

columns and the beams of the open first story are required to be increased by

increasing story shear and moment, by multiplying factor of 2.5. Then design for new

forces and again define plastic hinge properties in new section of soft story. After

nonlinear static analysis the new capacity curves shown in figure 4.2 to 4.7.

54

4.1.2 2.5 times increasing design forces in column only in soft story

In the case of the OGS frame strengthened using 2.5 times column & beams, it was

observed that plastic hinges developed in only the first-story columns. Stronger beams

would further increase the seismic demands on the columns. Therefore, increasing the

strength of first-story beams may exert additional force demands on the first-story

columns, whose strengths were also increased by using a predetermined multiplying

factor. This has been observed by Fardis and Panagiotakos 1997. As well, after

studying a similar clause in an older version of Eurocode 8 for buildings with severe

vertical irregularities. The new version of Eurocode 8 CEN 2003 requires that lateral

strength of only the first-story columns be increased. Also in Proposed Draft

Provisions and Commentary on Indian Seismic Code IS 1893 (Part 1) gives comment

on the clause for soft story that only column to be retrofitted by 2.5 times increasing

story shear and moment.

Figure 4.2 & 4.3 shows pushover curve i.e. capacity curve of G+3 building in X & Y

direction, when building is OGS then there is a formation of mechanism. It fails

before giving sufficient displacement or ductility. After retrofitting its stiffness,

strength increase as well as ductility. All form of hinges form and also energy

dissipated all over the frame model.

Figure 4.4 & 4.6 shows curve for G+7 & G+15 building in X-direction respectively.

After retrofitting it gives large deformation but it yields before as compare to OGS

model. Figure 4.5 & 4.7 shows curve in Y direction of G+7 & G+15 building. After

retrofitting strength, stiffness & roof displacement i.e., directly ductility increases.

0 0.02 0.04 0.06 0.08 0.1 0.12 0.14 0.16 0.180

2000

4000

6000

8000

10000

12000

14000

full infill

OGS

2.5 column

2.5 column & beam

Displacement (m)

Ba

se s

hea

r (

kN

)


55

Figure 4.2 Pushover curve of G+3 building retrofitting with 2.5 column & beam in X-

direction

0 0.05 0.1 0.15 0.2 0.25 0.30

2000

4000

6000

8000

10000

12000

Full infill

OGS

2.5 column only

2.5 column & beam

Displacement (m)

Ba

se s

hea

r (k

N)


Figure 4.3 Pushover curve of G+3 building retrofitting with 2.5 column & beam in Y-

direction

0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 0.4 0.450

3000

6000

9000

12000

15000

18000

21000

Full in-fill

OGS

2.5 RET of Col in X

2.5 col & beam in X

Displacement (m)

Bas

e sh

ear

(kN

)



direction

56

0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 0.40

2000

4000

6000

8000

10000

12000

14000

Full infill

OGS

2.5 Ret col

2.5 Ret col & beam

Displacement (m)

Bas

e sh

ear

( kN

)



direction

0 0.1 0.2 0.3 0.4 0.5 0.6 0.70

2000

4000

6000

8000

10000

12000

14000

16000

18000

20000

Full infill

OGS

2.5 Ret of col only

2.5 Ret of col & beam

Displacemnt (m)

Ba

se s

hea

r (k

N)



direction

57

0 0.1 0.2 0.3 0.4 0.5 0.60

1000

2000

3000

4000

5000

6000

7000

8000

9000

Full infill

OGS

2.5 Ret column

2.5 Ret column & beam

Displacementi (m)

Ba

se s

hea

r (k

N)



direction

4.1.3 Friction Damper

Retrofitting the structure with friction dampers provide the strength control required

to obtain the optimum structural response. Adding friction dampers increase stiffness

of the frame until a certain shear level is reached, at which the dampers can be set to

slip. For earthquake loading an appropriate slip level can be selected to give optimum

response. The energy dissipated by the friction dampers reduces the energy demand

on the structure and damps out the structural response. (Pall A.S. and Pall R. 1991)

Friction dampers are those that dissipate energy through the sliding of surfaces with

high coefficient of friction. Friction dampers are designed to have moving parts that

will slide over each other during a strong earthquake. When the parts slide over each

other, they create friction which uses some of the energy from the earthquake that

goes into the building. The earliest suggestion for the use of friction dampers is made

by W.O. Keightley in 1977. Keightley’s friction dampers are formed with two steel

plates slotted and clamped together with bolds and Belleville washers and lubricated

to prevent locking. The required normal force is provided by the tension in the bolts.

The Belleville washers are used to prevent loss of this tension. Energy is dissipated

after the tension or compression force applied to the plates along longitudinal

direction exceeds the friction force between the two plates and the plates slide one

58

with respect to the other. Application of cyclic load with a magnitude greater than the

slip force leads to rectangular force-deformation hysteresis loop.

Figure 4.8 Friction Damper in single diagonal (Pall A.S. and Pall R. 1996)

It is an effective way to control seismic response of structure and non-structural

damage. It does not impact the foundation design, features that make them

particularly attractive for upgrading existing buildings. They also seem to offer

savings in the initial cost of new structures and retrofitting of the existing ones

compared to conventional solutions. However, it is very difficult to maintain its

properties for long time intervals. It also has other disadvantages like they are

effective only for flexible structure, they encumber the design procedure and make it

more expensive and the selection of the appropriate slip load is a critical issue in the

performance of a structure with friction dampers.

In present work, friction dampers are defining as a link support element in which

Plastic (Wen) type damper are using. It is based on plastic wen theory which gives

elasto-plastic behaviour. It requires weight & some non-linear properties like stiffness,

yield strength, post yield stiffness ratio & yielding exponent.

Open ground story is specially constructed for parking. If we have to use friction

dampers in ground story then it disturbs vehicles to park. So by using trial & error

method best position of dampers such that it will not disturb parking such position&

also such yield strength were calculated. Only one direction i.e., in X direction & in

two internal frame of building friction dampers were used as it gives ease in parking

59

shown in figure 4.9. Figure 4.10 to 4.12 shows the capacity curve in X direction only

of G+3, G+7 & G+15 building using friction dampers.

After using friction dampers, capacity curve gives large displacement as compare to

open ground story means ductility increase and also increase in stiffness & strength

shown in figure 4.10 to 4.12. Table 4.1 shows comparisons of ductility in OGS &

retrofitting with friction dampers model. In G+3 model it increase from 1.6 to 4.24 &

in G+7 it increase from 2 to 5.86 & in G+15 it increase from 2 to 3.268.

Figure 4.9 G+3 building retrofitting with friction dampers showing Elevation & 3D view

0 0.02 0.04 0.06 0.08 0.1 0.12 0.14 0.160

2000

4000

6000

8000

10000

12000

14000

full infill

OGS

Ret with Fric damp.

Displacement (m)

Ba

se s

hea

r (

kN

)


Figure 4.10 Pushover curve of G+3 building retrofitting with Friction dampers in X-direction

60

0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 0.4 0.45 0.50

3000

6000

9000

12000

15000

18000

21000

Full in-fill

OGS

Ret friction

Displacement (m)

Ba

se s

hea

r (k

N)


Figure 4.11 Pushover curve of G+7 building retrofitting with Friction dampers in X-direction

0 0.1 0.2 0.3 0.4 0.5 0.60

3000

6000

9000

12000

15000

18000

21000

Full infill

OGS

Ret Frict dampers

Displacemnt (m)

Ba

se s

hea

r (

kN

)


Figure 4.12Pushover curve of G+15 building retrofitting with Friction dampers in X-

direction

Table 4.1 Ductility, yield force, ultimate force comparison in open ground story & retrofitting

with friction dampers model of different building in X- direction.

Type of building G+3 Ductility=

Δu/Δy G+7 Ductility= Δu/Δy G+15 Ductility=

Δu/Δy

Retrofitting with friction dampers

Fy = 9000kN

4.242

Fy = 15800kN

5.867

Fy=17100kN

3.268

Δy = 33mm

Δy =75mm

Δy=153mm

Fu = 10000kN

Fu = 18000kN

Fu =18800kN

Δu=140mm

Δu =440mm

Δu=500mm

Open Fy = 5400 1.600 Fy = 16633 2.000 Fy = 5400 2.000

61

ground story

kN kN kNΔy =30

mmΔy =80

mmΔy=125

mmFu = 5811

kNFu = 15200

kNFu = 5811

kNΔu =48

mmΔu=160

mmΔu=250

mm

4.1.4 Shear wall

Shear wall systems are one of the most commonly used lateral load resisting systems

in high rise buildings. Shear walls have very high in plane stiffness and strength,

which can be used to simultaneously resist large horizontal loads and support gravity

loads, making them quite advantageous in many structural engineering applications.

As a result of the large stiffness of walls, good story drift control can be achieved

during an Earthquake, as shown in Figure 4.13.

There are lots of literatures available to design and analyse of a shear wall. The

practice before 1960s has been to design buildings primarily for gravity loads and

check the adequacy of the structure for safety against lateral loads. It is established

that the design of a multi-storey building is governed by lateral loads and it should be

the prime concern of designer to provide adequate safety to structure against lateral

loads.

Figure 4.13 (a) Lateral load pattern; Lateral deformation pattern for (b) Frame element (c)

Wall element (d) Frame-wall building (Photo from Brama R.S. &Dasgupta K.)

Further, the old buildings were having substantial non-structural masonry walls,

partitions and connected staircases, which provided a significant safety margin against

lateral loads. The modern buildings are having light curtain walls, lightweight flexible

62

partitions along with high strength concrete and steel reinforcement, which reduce the

safety margin provided by non-structural components. Many existing RC frame

buildings located in seismic zones are deficient to withstand moderate to severe

earthquakes. Insufficient lateral resistances and poor detailing of reinforcement are the

main reasons for inadequate seismic performance of these structures. One of the most

popular methods of strengthening of seismically deficient structure is to provide shear

walls. “Shear walls” are defined as vertically oriented wide beams that carry

earthquake loads to the foundation. These are slender vertical cantilever RC walls

resisting the lateral load with or without frames. RC walls are often introduced into

multi-storey buildings at certain locations to resist lateral forces when frame systems

alone are insufficient. The term “shear wall” covers elevator shafts, stairwells and

central core units, in addition to plane walls. Shear walls acting with frames increase

the rigidity for lateral load resistance. When walls are situated in advantageous

position in a building, they can be very efficient in resisting lateral loads originating

from wind or earthquake.

Analysis for lateral loads of buildings containing shear walls is generally carried out

by assigning all lateral loads to the shear walls, since it is felt that the very big

difference in stiffness between the shear walls and the frame would cause the shear

walls to attract the total lateral loads. Shear walls in high seismic regions require

special detailing. However, in past earthquakes, even buildings with sufficient amount

of walls that were not specially detailed for seismic performance (but had enough

well-distributed reinforcement) were saved from collapse. Shear walls are easy to

construct because reinforcement detailing of walls is relatively straight-forward and

therefore easily implemented at site. The general method of providing a shear wall is

to fill the gaps between columns of the moment resisting frame with partial or

complete reinforced concrete wall is termed as Internal Shear Wall (Buttress Wall). A

new concept of providing walls on the outside is now being adopted in practice

(Yasaret al. 2008) known as External Shear Walls, Placing of reinforced concrete wall

along the external face of column with or without coupling beams is termed as

External Shear Wall. These types of shear walls are said to be advantageous

considering the fact that the building usage is not disturbed during retrofit. A study

has been carried out by M. Ashraf on optimum location of shear wall in a multi-storey

building. Few investigations have been carried out to evaluate the efficiency of

63

external shear walls in comparison to internal shear walls and to find the advantages

imparted by external shear walls to bare frame when subjected to lateral (Earthquake)

load. Present paper focuses on the influence of shear wall compared with that of

without shear wall on the resulting forces of the Moment Resisting Bare Frame under

consideration with different orientation and location of shear wall when subjected to

same type of load. A total of 5 models subjected to lateral load that may arise due to

earthquake on a fifteen storied building in zone II of the Seismic zones of India were

considered for the study. The top storey sway, support reaction, Column forces,

bending moment between the structural systems were compared with each other.

4.1.4.1 Addition of shear wall

Addition of shear wall into an existing building is most common approach of seismic

retrofitting. It has been used with frame, since long time. It is an effective method of

increasing building strength and stiffness. When shear walls are situated at proper

positions in a building, they can form an efficient lateral-force resisting system, while

simultaneously fulfilling functional requirements. Addition of shear wall improves

buildings strength and stiffness, and also it is economically feasible and readily

compatible with most of existing concrete buildings. It also gives good aesthetic view.

4.1.4.2 Modelling of shear wall

The analytical model for a solid wall element should represent the strength, stiffness

and deformation capacity of the wall for inplane loading. Out of plane behaviour need

not be consider, except where the wall acts as a flange for an intersecting wall

element. Solid walls may be considered “slender” if their aspect ratio (height/length)

is equal to or exceed 4 (hw/tw>= 4). Solid wall may be considered “squat” if their

aspect ratio is less than or equal to 2 (hw/tw =< 2). Slender walls usually are controlled

by flexural behavior, although shear strength may be a limiting factor in some cases.

Squat walls usually are controlled by shear behavior, although flexure sometimes may

be a limiting factor. The response of walls with immediate aspect ratios usually is

influenced by both flexure and shear. Potential failure of anchorages and splices may

require modelling of these aspects as well. Interaction with other elements should be

64

represented. Except for squat one and two story walls, sliding along construction

joints need not be modelled (ATC-40).

Various analytical models have been proposed in literature to simulate the behaviour

of RC shear wall. These include Equivalent beam model, wide column model, thin

shell model, three vertical line element model, inelastic analytical macroscopic model,

Macro-finite-element model.

In equivalent beam model the shear wall member is replaced at its centroidal axis by a

line element and connected by rigid link to the frame beams. The main limitation of

this model lies in the assumption that rotation occurs around points belonging to the

centroidal axis of the wall so it does not accounts for migration of the neutral axis of

the wall cross-section, rocking of the wall etc. In three vertical line element model, a

generic wall member is idealized as three vertical line element with infinitely rigid

beam at the top and bottom floor levels. Two outside truss elements represent the axial

stiffness of the boundary columns, while the central element was one-component

model with vertical, horizontal and rotational springs concentrated at the base. This

model is capable of describing flexural and shears deformation including the

migration of the neutral axis of the wall cross-section, rocking of wall etc., but the

deformation due to the fixed end rotation and web splitting-crushing mode of failure

is not accounted [Vulcano and Bertero 1987]. Inelastic analytical macroscopic model

uses eight inelastic axial springs connected by two rigid beams to account plastic

bending deformation of wall, and three shear springs which expresses the shear

behaviour of panel and two boundary columns [Fu,et al., 1992]. Macro-Finite-

Element model consists of a number of vertical elements. These vertical elements

consist of vertical and horizontal springs at the center of each vertical element. The

axial stiffness of each vertical spring is represented by two parallel components

representing mechanical behaviour of the concrete and steel. Horizontal spring

represents the shear springs. The stiffness of each shear spring is determined by the

different state of vertical spring. In this model the axial springs first reach the

nonlinear state and then the shear spring, thus the effect of axial stiffness on shear

stiffness was neglected.

In present work, shear wall has been model using equivalent wide column modelling

method. The shear wall member is replaced at its centroidal axis by a line element and

connected by rigid link to the frame beams. Line element is defining as column whose

65

width and length is shear wall’s width and length. Column width and length has been

taken as 0.23 m and 4.6 m respectively. Height to length (hw/tw) ratio is 0.78 means

shear wall is squat. Squat walls usually are controlled by shear behaviour. The wide-

column analogy was originally developed for planar wall structures such as walls with

openings (e.g.Clough et al., 1964) and was later extended to non-planar structures

(e.g. MacLeod and Hosny, 1977). Despite being widely applied for seismic analysis of

structures, only very little literature on wide-column models within elastic properties

has been found. Most of the research carried out in the past concentrated on the

behavior of wide-column models with elastic properties and on eliminating any

disadvantages related to the discretization of the walls into beam elements.

4.1.4.3 Analysis and design

In present work, shear wall has been model as wide column modelling as in non-

linear static analysis we requires non-linear properties and also define hinges of walls.

So it’s easy to give auto hinges to column as P-M2-M3. To check whether wide

column modelling is correct or not also thin shell model of shear wall has been

modelled. After design in both cases it gives approximately same percentage of steel.

Figure 4.15 and 4.16 shows percentage of steel in wide column modelling and thin

shell modelling in G+3 building respectively.

66

For optimization of best position of shear wall trial and error method has been used

and calculate best position. Then Shear wall is assigned in alternate position in x

direction i.e., longitudinal direction as shown in figure 4.17. Position is such that it

gives large displacement in capacity curve.

Figure 4.15. Thin shell model showing percentage of steel after addition of shear wall.

67

Figure 4.16 G+3 building retrofitting with shear wall showing 3D view of position of shear

wall

Figure 4.17 Pushover curve of G+3 building retrofitting with shear wall in X-direction

68

0 0.05 0.1 0.15 0.20

2000

4000

6000

8000

10000

12000

14000

full infill

OGS

Shear wall

Displacement in m

Ba

se s

hea

r in

kN

B- IO- LS CP - DBE- MCE-

Figure 4.18 Pushover curve of G+3 building retrofitting with shear wall in Y-direction




69

0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 0.40

3000

6000

9000

12000

15000

18000

21000

Full infill

OGS

Ret Shear wall

Displacement in m

Ba

se s

hea

r k

N

0 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.80

2000

4000

6000

8000

10000

12000

14000

16000

18000

20000

Full infill

OGS

Ret. Shear wall in-X

Displacemnt in m

Bas

e sh

ear

in k

N

0 0.05 0.1 0.15 0.20

1000

2000

3000

4000

5000

6000

7000

8000

9000

10000

Full infill

OGS

Shear wall in y

Displacement in m

Ba

se s

hea

r in

kN

0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 0.40

2000

4000

6000

8000

10000

12000

Full infill

OGS

Ret. Shear wall in Y

Displacement in m

Ba

se s

hea

r i

n k

N


Figure 4.17 shows pushover curve of G+3 building in x direction. In which

comparison of capacity curve of full infilled, open ground story and retrofitted model

of shear wall. When retrofitting has been done it shows large deformation (figure

4.17). Initial stiffness also increased. Figure 4.18 shows pushover curve in Y direction

of G+3 building as there is no requirement of retrofitting in Y direction but due to

addition of walls in X direction it also affects in Y direction. Its initial stiffness

changes, load carrying capacity changes. But its analysis stops before due to

formation of mechanism and ductility reduced.

Figure 4.19 also shows pushover curve of G+7 story model in X direction with 3

different model but in this case ultimate load carrying capacity of retrofitted model is

decreases as compare to open ground story but ultimate displacement is drastically

increase. And figure 4.20 shows curves in Y direction same condition as described for

figure 4.18.

70

0 0.1 0.2 0.3 0.40

1000

2000

3000

4000

5000

6000

7000

Full infill

OGS

Ret. Shear wall in-Y

Displacementi in m

Bas

e sh

ear

in k

N

Figure 4.21 shows pushover curve of G+15 building in X direction with 3 different

models. In this case load carrying capacity, ultimate displacement, and initial stiffness

has been increased. Also in Y direction in figure 4.22 all three properties i.e., initial

stiffness, load carrying capacity, and ultimate displacement increase compare to open

ground story model.

Figure 4.24 G+3 building retrofitted with shear wall in X-direction showing hinges

formation.

Figure 4.23 shows different hinges formation in G+3 building retrofitted with shear

wall. There is no deformation in ground story compare to above stories. And analysis

stops due to formation of mechanism. In first story, in all columns hinges formed in

lower level.

After using shear wall, capacity curve gives large displacement as compare to open

ground story means ductility increase and also increase in stiffness & strength shown

in figure 4.17. Table 4.2 shows comparisons of ductility in OGS & retrofitting with

shear wall model. In G+3 model it increase from 1.6 to 7.118& in G+7 it increase

from 2 to 7.174& in G+15 it increase from 2 to 7.07.

Table 4.2 Comparison of ductility, yield force & ultimate force in open ground story &

retrofitting with shear wall model of different building in X- direction.

Type of building G+3 Ductility=

Δy/Δu G+7 Ductility= Δy/Δu G+15 Ductility=

Δy/Δu

Retrofitting Fy = 11105 7.118 Fy = 14047 7.174 Fy=17389 7.071

71

with Shear wall

Δy =17mm Δy = 46mm Δy=113mm

Fu = 12647 Fu = 15403 Fu =18964

Δu=121mm Δu =330mm Δu=799mm

Open ground story

Fy = 5400

1.600

Fy = 16633

2.000

Fy = 5400

2.000Δy =30mm Δy = 80mm Δy=125mm

Fu = 5811 Fu = 15200 Fu = 5811

Δu = 48mm Δu =160mm Δu=250mm

Chapter 5

CONCLUSIONS AND SCOPE FOR FUTURE WORK

5.1 Conclusions drawn from work

In India, most of the existing as well as new infilled RC frame buildings has been

designed and are being designed without considering strength and stiffness of Infills

(bare frame modeling). Due to inclusion of infills, behavior and failure modes of

buildings changes. This leads to serious concern about seismic safety of existing

buildings. In the present study, various strengthening techniques for open ground story

72

have been discussed. These techniques can be broadly categorized in two groups; 1

Strengthening of existing members, 2 Addition of new members.

1) In the first group, there are two methods; reinforced concrete jacketing and

steel jacketing. In the present work, jacketing is done by

a) Redesign beam and column of open ground story by increasing 2.5 times

seismic design forces. In present work, due to jacketing it gives good

ductility and increase up to 3 times compare to open ground story and also

increase in strength carrying capacity and initial stiffness of open ground

story.

b) Redesign only column of open ground story by increasing 2.5 times

seismic design forces. Also in this method it gives good ductility it

increase up to 3.5 times compare to open ground story and also increase in

strength carrying capacity and initial stiffness of open ground story.

2) In the second group, there are two popular methods; addition of shear wall and

addition of friction dampers.

a) Addition of friction damper is attractive and easy to construct but needs

sophisticated method for proper fixation with existing frames. In present

work, due to retrofitting with friction damper, there are change in ductility

it increase up to 2.5 times compare to open ground story and also increase

in strength carrying capacity and initial stiffness of open ground story. For

Indian RC framed buildings, it is observed that beam column joints are

most vulnerable to seismic failure, sometimes; addition of dampers

without proper strengthening of joints may lead to catastrophic failure of

building.

b) Addition of new shear wall can efficiently be used for buildings with only

local interventions. In present work, due to addition of shear wall

ductility which is increase up to 3.5 times compare to open ground story

and also increase in strength carrying capacity and initial stiffness of open

ground story. On the other hand, addition of shear wall needs laying of

new foundation, which in itself a difficult task.

In many cases, it could be difficult to achieve a single retrofitting technique for

attaining the desired performance of buildings. Combination of some of the above

mentioned techniques may be required. However, to determine the performance of

73

these techniques both experimental as well as analytical verifications in conjunction

with Indian construction practice are needed.

The RC frame with open-ground-storey exhibited very poor lateral strength stiffness

and energy dissipation capacity due to formation of shear hinges in ground-storey

columns under lateral load resulting uncontrolled excessive deformation in the

ground-storey in a nonlinear static analysis of a typical building.

5.2 Future work

1. Single strut model for infills can accurately predict the lateral stiffness and

strength of masonary infilled RC frame. However, use of single strut can only

take into account its compressive failure; it can’t predict local failure in frame

member. Single strut models underestimate the force resultants in frame

member.

2. In the present study, openings were not considered in infills. Presence of

opening in infills significantly reduces the stiffness and strength of the infilled

frames. Suitability of the proposed strengthening schemes must be verified for

masonary-infilled frames with openings with walls.

3. Also for future work, non linear dynamic analysis (time history analysis) is a

best method for analyzing the strengthening methods like friction dampers.

4. The experimental work should be carried out on a reduced scale three story

with first story without infilled wall under gradually increased cyclic lateral

displacements to further verify the effectiveness of proposed strengthening

schemes.

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