ETGE2012_proceedings.pdf

Emerging Trends in Geotechnical Engineering

Proceedings of National workshop on

EEmmeerrggiinngg TTrreennddss iinn GGeeootteecchhnniiccaall EEnnggiinneeeerriinngg ((EETTGGEE 22001122)) 8th June 2012, IIT Guwahati, Guwahati

Edited by A. Murali Krishna

Department of Civil Engineering Indian Institute of Technology Guwahati

and Indian Geotechnical Society Guwahati Chapter (NE)

Guwahati

Organised by

Proceedings of National Workshop on Emerging Trends in Geotechnical Engineering (ETGE 2012), Guwahati, 08th June 2012

Emerging Trends in Geotechnical Engineering Edited by A. Murali Krishna Department of Civil Engineering Indian Institute of Technology Guwahati Guwahati. Workshop Organisers: Department of Civil Engineering Indian Institute of Technology Guwahati Indian Geotechnical Society Guwahati Chapter (NE) Guwahati

National Workshop on Emerging Trends in Geotechnical Engineering (ETGE 2012), Guwahati, 08th June 2012 Organising Committee: Chairmen: Prof. AK Sarma, Department of Civil Engineering, IIT Guwahati Prof. UC Kalita, President, IGS Guwahati Chapter (NE) Organising Secretaries: Dr. A. Murali Krishna, IIT Guwahati Dr. Diganta Goswami, Assam Engineering College, Guwahati Dr. Utpal Barua, Assam Engineering Institute, Guwahati Members: Shri. KL Das, Guwahati Dr. Baleshwar Singh, IIT Guwahati Dr. S. Sreedeep, IIT Guwahati Dr. Arindam Dey, IIT Guwahati Dr. Binu Sharma, Assam Engineering College, Guwahati Mr. Sashanka Bora, Assam Engineering College, Guwahati Mr. Abinash Mahanta, Assam Engineering College, Guwahati Dr. Kumar Pallav, IIT Guwahati Sponsors: Purbanchal Cements Ltd. Indian Geotechnical Society, New Delhi.

Workshop on Emerging Trends in Geotechnical Engineering (ETGE 2012) 8th June 2012, Guwahati

Preface India has emerged as one of the fastest growing economies in the world. Various new infrastructure projects are under progress across the country especially in Northeast India. Geotechnical engineering is the first and most important aspect of any infra project and urban development. Over the years many new trends were being developed and practiced in terms of analysis, design and construction of geotechnical structures and associated site investigations. In this vein, a National workshop on “Emerging trends in Geotechnical Engineering” (ETGE 2012) is organised under the auspices of the Department of Civil Engineering, Indian Institute of Technology Guwahati in association with Indian Geotechnical Society Guwahati Chapter (NE). I express sincere gratitude to both the organisations for their support. “Emerging trends in Geotechnical Engineering”, encompass many different subdivisions of the field of Geotechnical Engineering. To name: Trends in Site investigation; Trends in laboratory testing, analysis; New trends in design methods and designs; Trends in ground improvement methods; Trends in geoenvironmental applications; Trends in utilisation of waste by products; Trends in predicting soil behaviour; Trends in dealing with underground structures offshore geotechnical structures; Trends in earthquake related problems and designs; Trends in construction methods under challenging situations etc. North eastern states being the most rapid developing parts of India, in terms of various infrastructure projects and hydro projects, are the places where the application of these new trends with amalgamation of classical soil mechanics principles is vital. There is a need for students, academicians and practicing professionals for updating their knowledge about the emerging trends in various geotechnical areas. This workshop presents some of the emerging trends in the geotechnical engineering practice and research activities. I am very grateful to the speakers for accepting my request to delivering the lectures and preparing the paper contributions. It gives me great pleasure to bring out the workshop proceedings. In this edited volume, about a dozen contributions are included from the eminent geotechnical researchers in India and other academicians, practicing

engineers. The chapters of this book covers and presents some of the ongoing trends in analyses of ground improvement methods, utilisation of waste products, ground stabilisations, ground response analyses of rock structures, foundation treatments, designs and analysis, geoenvironmental challenges and applications, along with construction trends for roads under challenging conditions. I am very thankful to all the authors for their time and effort in developing the workshop lectures and sharing their technical experiences. I hope the proceedings will enrich knowledge of the workshop participants and helps for the future growth of geotechnical engineering with new trends. I would to express my sincere thanks to the organising committee for their support in organising this workshop. On behalf of organising committee, I extend our sincere thanks to the sponsors Purbanchal Cements Ltd., Guwahati and Indian Geotechnical Society, New Delhi. I also thank all the student volunteers for their support and assistance in all respects of smooth conduct of the workshop. Last, but not the least I express thanks for all the participants to making this workshop ETGE 2012, a grand success. IIT Guwahati Murali Krishna

June 2012

Workshop on Emerging Trends in Geotechnical Engineering (ETGE 2012) 8th June 2012, Guwahati.

Contents Sl. No. Title Page No.

1. Soft Ground Improvement with PVDs --- M. R. Madhav and Ayub Khan

1

2. Coal Ashes in Geotechnical Engineering Practice: Beneficial Aspects

--- A. Sridharan

11

3. Ground Response and Support Measures for a Railway Tunnel in the Himalayas

--- K. S. Rao

27

4. Treatment of Foundations and Geological Faults of Almatti Dam on Krishna River: A Case Study

--- M. Bidasaria

37

5. Recent Experiences of Ground Stabilization Techniques --- Satyendra Mittal

45

6. Performance Based Earthquake Resistant Design of Geotechnical Structures – A New Trend

--- S. K. Prasad and P. Nanjundaswamy

61

7. A Decision Support System for Risk Assessment and Remediation Option Selection for Contaminated Soils and Groundwater

--- R. K. Srivastava

75

8. Risk Analysis of Bearing Capacity of Shallow Foundations --- Dasaka S Murty

89

9. Prediction of Soil Behavior – A Reappraisal --- Binu sharma

99

10. Road Embankments in Water Logged and Frost Affected Areas – Problems and Solutions

--- Jai Bhagwan and Kanwar Singh

113

11. The Principles and Application of Geo-Environmental Engineering

--- Anil Kumar Mishra

119


1

Soft Ground Improvement with PVDs

M.R. Madhav1 and Ayub Khan, P.2

1Professor Emeritus, JNTUH CE and Visiting Professor, IIT, Hyderabad, email: [email protected]

2ACE Engineering College, Ankushapur, Ghatkesar, email: [email protected] ABSTRACT: Treating soft ground with PVDs is one of the most popular techniques for improving soft soil deposits. PVDs improve the ground by accelerating consolidation arising from preload. This paper presents an analysis of non-linear theory of radial consolidation due to PVDs in thick soft clay deposits. Keywords: PVDs, Ground improvement, soft ground, consolidation, non-linear theory

1. INTRODUCTION India has a long coast line of nearly 5,000 km length. Soft, weak, highly compressible soft soils are prevalent along this stretch. Considerable infrastructure development is taking place along the coast for obvious reasons. Construction of facilities on these deposits is very challenging because of their low strength, high compressibility, etc. Several alternatives, e.g. preloading without or with PVDs, granular piles, heavy tamping, etc., are available to engineer the ground. Amongst these, preloading with PVDs and reinforcing with granular piles/stone columns are the most preferred alternatives.

2. PREFABRICATED VERTICAL DRAINS (PVDs) These are band or strip shaped plastic drains about 100 mm wide and 4 mm thick usually installed in square or triangular arrays (Fig. 1a). Consequently, the zone of influence of each drain is either square or hexagonal area. The strip drain and the zone of influence of each drain are replaced by a unit cell of equivalent circular shapes. The flow pattern around the drain is considered to be axi-symmetric (Fig. 1b). The equivalent diameter of the drain, dw =2(a+b)/p, where ‘a’ and ‘b’ are the width and thickness of the PVD respectively and the equivalent diameter of the zone of influence, de = 1.13S & 1.05S for square and triangular patterns respectively, where S is the spacing of drains. Barron (1947) presented analytical solutions for the radial consolidation due to sand drains for both free and equal strain conditions. This classical theory is based on the assumptions of small strains, linear void ratio-effective stress relationship and constant coefficients of volume compressibility, mv, and horizontal permeability kh. Hansbo (1981) presented a simple solution for radial consolidation with band shaped vertical drains. Lekha et al. (1998) presented a non-linear theory of consolidation with sand drains under time dependent loading for equal strain case. Full-scale test conducted by Bergado et al. (2002) on soft Bangkok clay with PVDs revealed that the degree of consolidation obtained from pore pressure measurements is lower than the corresponding values obtained from settlement measurements.

2 ETGE 2012

de 1

1

S

S

(a) (b)

dw

H Drain

de

Figure 1(a) Triangular Arrangement of PVDs and (b) Flow in Unit Cell The various modeling aspects of PVDs are comprehensively discussed by Indraratna et al. (2003) along with the evaluation of their effectiveness in practice. Indraratna et al. (2005a) developed a theory for consolidation with radial flow using s ¢log - e (Cc and Cr) & e – log kh (Ck) relationships and for different loading increment ratios )/( iss ¢¢D . Indraratna et al. (2005b) developed a modified consolidation theory for vertical drains incorporating vacuum preloading for both axi-symmetric and plane strain conditions. Two- and three-dimensional multidrain finite-element analyses of a case study of combined vacuum and surcharge preloading with vertical drains is presented by Rujikiatkamjorn et al. (2008). The numerical predictions compared well with observed data. Indraratna et al. (2008) developed a new technique to model consolidation by vertical drains beneath a circular loaded area by transforming system of vertical drains into a series of concentric cylindrical drain wall. Walker et al. (2009) presented the spectral method for analysis of vertical and radial consolidation in multilayered soil with PVDs by assuming constant soil properties within each layer. For relatively large applied stress range the void ratio is not proportional to effective stress and the coefficients of compressibility and permeability decrease during consolidation. A non-linear theory of consolidation was developed by Davis & Raymond (1965) considering e – log σ΄ relationship but it is valid only for vertical flow and thin layer of clay. A theory of non-linear consolidation for radial flow around a vertical drain is developed by Ayub Khan et al. (2010a) based on linear void ratio-log effective stress relationship but assuming constant coefficient of consolidation for thin clay layers. This theory is further extended by Ayub Khan et al. (2010b) for thick clay deposits as well. The general equation of non-linear consolidation with radial flow (Ayub Khan et al. 2010b) in terms of a parameter, w, is

÷÷ø

öççè

æ

¶¶

+¶

¶=

¶¶

rw

rrwc

tw

r1

2

2 (1)

with f

10logwss

¢¢

= or f

f10

ulogw

ss

¢-¢

=)( , where σ΄ and u are the effective vertical stress and the

excess pore pressures respectively at time, t and radial distance, r, cr - coefficient of consolidation for radial flow and σ΄f the final effective vertical stress. The parameter ‘w’ varies with the depth in thick deposits of clay as the initial effective in-situ stress, σ΄o and the final effective stress, σ΄f (=σ΄o +qo) vary with depth due to overburden stresses, and qo is the applied load intensity. The thick clay layer of thickness, H is divided into m layers of

Soft Ground Improvement with PVDs 3

thickness, ΔH=H/m (Fig. 2). The equation governing the consolidation process of each layer is

÷÷÷

ø

ö

ççç

è

æ

¶

¶+

¶

¶=

¶

¶

rw

rr

wc

tw jj

rj 1

2

2 (2)

where the subscript j refers to the layer number and wj = wj (r, t) of the jth layer. Eqs. (1 and 2) are of the same form as that given by Barron (1947). Even though the initial and final effective stresses are different in each layer, but the flow in each layer is assumed to be purely radial and independent of the flows in the adjacent layers. Degree of Settlement of the layer j is,

÷÷÷÷÷

ø

ö

ççççç

è

æ

ò

ò

-=

ò -

ò -

=e

w

e

w

e

w

e

wr

rjo

r

rj

r

rfo

r

ro

jS

drrw

drrw

drree

drree

U 2

2

1 2 )(

2 )(

,

,

p

p

p

p

(3)

where jf

j10j logw

,s

s¢

¢= ,

jf

jo10jo logw

,

,, s

s¢

¢= , the initial

effective in-situ stress in the jth layer, σ΄o,j .zj.

Centre Line of Drain

Layer- m

H/m

Surcharge, qo Drain

Layer-1

rw re

o f

Impervious

Pervious

z Flow Path

Figure 2 Radial Flow in Clay Layers in a Thick Deposit

Average degree of consolidation for the entire thickness,

H

HUU

m

jjjs

s

å D

= =1, ).(

(4)

Normalized average excess pore pressure for the layer j is:

ò

ò=

e

w

e

wr

ro

r

rj

javgdrru

drru

U 2

2

* ,p

p

(5)

4 ETGE 2012

Degree of dissipation of average excess pore pressure at the layer, j is:

)*1( ,, javgjP UU -= (6) Average degree of dissipation of excess pore pressure for the entire thickness is:

H

HUU

m

jjjP

P

å D

= =1, ).(

(7)

Initial and Boundary Conditions

jf

jo10joj

jojfj

logrwrw

ru

,

,,

,,ew

)0,()0,(or

)()0,( ; rr r and 0For t

s

s

ss

¢

¢==

¢-¢=££=

(8) where σ΄f,j= (σ΄o,j+Δσ΄) and Δσ΄=qo

0),(or 0),( ; rr and 0For t w =====> trrwtrru wjwj (9)

0or 0 ; and 0For t =¶

¶=

¶

¶=>

== ee rr

j

rr

j e r

wr

urr

(j=1, 2, 3…m) (10) The radius of influence zone, re = 0.5de and radius of equivalent drain, rw =0.5dw. Eq. (2) is re-written in non-dimensional form as

÷÷

ø

ö

çç

è

æ

¶

¶+

¶

¶=

¶

¶

Rw

RR

wT

w jjj 12

2

( j=1, 2, 3……m) (11) where R = r/de and time factor, Th = cr.t./de

2. 3. RESULTS AND DISCUSSION A thick clay layer of thickness, H, is divided into ‘m’ (=20) layers of equal thickness, ΔH= H/m, as shown in Figure 2. The numerical analysis is carried out for each layer independently using the corresponding ratio σ΄f/σ΄o and the results obtained in terms of degree of settlement, Us, and excess pore pressures, u. Variations of these results along the depth and radial distance are studied. As non-linear consolidation is mainly influenced by the stress ratio, σ΄f /σ΄o, the variation of σ΄f/σ΄o with depth is shown in Fig. 3 for different values of normalized applied load intensity, q*o=qo/(γ΄.H). The decrease of σ΄f /σ΄o with depth is very sharp as the initial effective stress is very small near the top. The decrease of σ΄f /σ΄o with depth is significant for depths in the range 0.1H to 0.4H and relatively very small for z>0.4H for all q*o. Hence, the effect of non-linearity in the void ratio - effective stress on consolidation in a thick clay layer can be pronounced at shallow depths compared to that at greater depths. Increases in q*o can be either due to increase of load intensity, qo for a given thickness, H, or due to decrease of thickness of clay deposit for a given load intensity. In either case, only the non-dimensional stress ratio, σ΄f/σ΄o, influences the rate of consolidation and not the thickness of the deposit or the magnitude of loading individually. The degree of settlement, Us, is determined for different layers along the depth for various qo* values for different values of n ranging from 5 to 40 and shown in Fig.4. The degree of


settlement is identical at all the layers in given thick clay for all values of n and q*o. The degree of settlement obtained from the present non-linear theory is identical to that from the linear theory (Barron 1947) for free strain case. The degree of settlement, Us, is independent of σ΄f /σ΄o and varies only with n as is the case for vertical flow (Davis and Raymond 1965). The degree of settlement decreases with increase of n since it takes relatively long time for dissipation of pore pressure for larger radial distances. Us decreases from 58% to 24.8% for n increasing from 5 to 40 at a time factor of 0.10.

Figure 3 Variation of σf΄/σo΄ with Depth

Figure 4 Us versus Th for both Linear and Non-Linear Theories – Effect of ‘n’

6 ETGE 2012

The degree of dissipation of average excess pore pressure for the entire thickness, Up is presented in Fig. 5 for n=15 along with the degree of settlement, Us. While the degree of settlement is independent of q*o, the degree of dissipation of pore pressure, Up is sensitive to q*o values. Thus for a given thickness of clay deposit, the increase of load intensity, qo results in decrease of degree of dissipation of average excess pore pressure or for a given load intensity the decrease of thickness of clay deposit results in decrease of degree of dissipation of average excess pore pressure.

Figure 5 Variations of Us and Up with Th

The normalized average excess pore pressure, U*avg(z) = (uavg(z)/uo).100 at different depths determined and its variation with time shown in Fig. 6 for n=15 and q*o=1 along with the degree of settlement. The excess pore pressure is relatively high at shallow depths where the σ΄f /σ΄o ratio is relatively high compared to the values at greater depths. The importance of the above finding is that when large load is applied on soft ground, the possibility of shallow seated rotational bearing failure is to be examined in view of the large residual pore pressures at these depths over longer periods of time. The normalized average excess pore pressures, U*avg (z) at different depths of the thick clay are presented in Fig. 7 for n=15 at time factor, Th=0.20 along with the results of linear theory. The remarkable phenomenon observed is that average pore pressure values from the non-linear radial consolidation theory vary with depth in contrast to the depth-independent U*avg values of linear theory. The differences between the pore pressures of non-linear and linear theories are relatively large at shallow depths due to large values of σ΄f /σ΄o compared to those at greater depths. This difference increases with increase of q*o. Moreover, at shallow depths the variation of pore pressure with depth is relatively very large compared to that at greater depths due to sharp variation of σ΄f /σ΄o at shallow depths. While the excess pore pressures in the linear theory are independent of, q*o, the pore pressures according to non-linear theory are dependent on q*o as the variation of q*o influences the ratio σ΄f /σ΄o..


The residual average excess pore pressures thus are underestimated in the conventional linear theory. In view of the above, instead of applying the entire preload instantaneously on the soft ground, it may be applied in increments with proper time lag to allow quicker dissipation of pore pressures and gain of shear strength.

Figure 6 Variation of Us & U*avg(z) with Th –- Effect of Depth

Figure 7 Variation of U*avg(z) with Depth—Effect of Load Intensity

Fig. 8 shows the pore pressure variation with radial distance at different depths for q*o=1, n=15 and Th=0.20. The excess pore pressures are relatively large at shallow depths wherein the stress ratio is extremely large but decrease with depth since the ratio of final to initial stress decreases with increase of depth. The pore pressure variation with radial distance, r, is relatively more significant in the upper half of the deposit compared to that in the lower half.

8 ETGE 2012

Figure 8 Variations of Pore Pressures with Radial Distance Effect of Depth

4. CONCLUSIONS Analysis of consolidation of soft ground treated by PVDs is presented. A simple non-linear theory of radial consolidation developed for thick deposit of clay treated with PVDs considering linear void ratio-log effective stress relationship predicts that while the degree of settlement is independent of the final to initial stress ratio, the degree of dissipation of pore pressure is very much dependent on the stress ratio. The residual excess pore pressures are under-estimated in the conventional linear theory. The proposed nonlinear theory substantiates the actual in-situ slower rate of degree of dissipation of excess pore pressures compared to that of the degree of settlement. The non-linear consolidation effect is pronounced at shallow depths compared to the effect at greater depths. The excess pore pressures due to radial drainage vary not only with time and radial distance but also with depth in contrast to depth-independent pore pressures from the conventional theory for radial flow. The significance of the proposed theory is that it can explain failure of high embankments constructed rapidly on thick deposits of fine grained soils.

REFERENCES Ayub Khan, P., Madhav, M.R. and Saibaba Reddy, E. (2010a). Effect of non-linear

consolidation for radial flow on pore pressure dissipation. Indian Geotechnical Journal, 40(1), 47-54.

Ayub Khan, P., Madhav, M.R. and Saibaba Reddy, E. (2010b). Consolidation of thick clay layer by radial flow-nonlinear theory. Geomechanics and Engineering, 2 (2), 157-160.

Barron, R.A. (1947). Consolidation of fine grained soils by drain wells. Trans. ASCE, 113(2346), 718-754.

Bergado, D.T., Balasubramaniam, A.S., Fannin, R.J. and Holtz, R.D. (2002). Prefabricated vertical drains (PVDs) in soft Bangkok clay: a case study of the new Bangkok International Airport project. Can. Geotech. J., 39, 304-315.

Davis, E.H. and Raymond, G.P. (1965). A non–linear theory of consolidation. Geotechnique, 15(2), 161–173.

Hansbo, S. (1981). Consolidation of fine grained soils by prefabricated drains. Proc. of the


10th ICSMFE, Stockholm, Sweden, June, 3, 667-682. Indraratna, B., Aljorany, A. and Rujikiatkamjorn, C. (2008). Analytical and numerical

modeling of consolidation by vertical drains beneath a circular embankment. Intl. Jl. of Geomechanics, 8 (3), 119–206.

Indraratna, B., Bamunawita, C., Redana, I.W. and McIntosh, G. (2003). Modeling of prefabricated vertical drains in soft clay and evaluation of their effectiveness in practice. Ground Improvement, 7(3), 127–137.

Indraratna, B., Rujikiatkamjorn, C. and Sathananthan, I. (2005a). Radial consolidation of clay using compressibility indices and varying horizontal permeability. Can. Geotech. J., 42(5), 1330–1341.

Indraratna, B., Sathananthan, I., Rujikiatkamjorn, C. and Balasubramaniam, A.S. (2005b), Analytical and numerical modeling of soft soil stabilized by prefabricated vertical drains incorporating vacuum preloading. Intl. Jl. of Geomechanics, 5(2)114–124.

Lekha, K.R., Krishnaswamy, N.R. and Basak, P. (1998). Consolidation of clay by sand drains under time–dependent loading. J of Geotech. and Geoenv., 124(1), 91-94.

Rujikiatkamjorn, C., Indraratna, B. and Chu, J. (2008). 2D and 3D numerical modeling of combined surcharge and vacuum preloading with vertical drains. Intl. Jl. of Geomechanics, 8(2), 114–156.

Walker, R., Indraratna, B. and Sivakugan, N. (2009). Vertical and radial consolidation analysis of multilayered soil using the spectral method, J. of Geotech. And Geoenv.Engrg, ASCE, 135(5), 657– 663.

10 ETGE 2012


11

Coal Ashes in Geotechnical Engineering Practice: Beneficial Aspects

A. Sridharan

INSA Honorary Scientist, Formerly Professor of Civil Engineering, Indian Institute of Science, Bangalore, email: [email protected]

ABSTRACT: Coal ashes have been shown to have advantageous properties such as low specific gravity, lower compressibility, higher rate of consolidation, high strength, high CBR, high volume stability, water insensitiveness to compaction and pozzolanic reactivity. The use of coal ashes having these beneficial properties, which are being considered as industrial wastes, serves as a very use full material in the field of geotechnical engineering. Their use in bulk in the field of geotechnical engineering is an eco-friendly way of their safe disposal. Keywords: coal ash, compressibility, consolidation, compaction, CBR

1. INTRODUCTION Since, the economic development of any country is directly related with the energy production and consumption of that country, more thrust is being applied of late on the electrical power generation sector. Due to their inherent limitations, establishing either the large scale hydroelectric power plants or nuclear power plants is receiving lesser priority. Instead, installing the coal based thermal power plants is being encouraged worldwide. The burning of pulverised coal in thermal power plants results in the production of huge quantum of coal ashes namely fly ash and bottom ash. The wet disposal of these ashes separately or in combination in storage ponds results in pond ashes. With the depletion of high quality coal resources, low quality coals are also being used, which enhance the quantum of coal ashes generated. The current worldwide production of coal ashes is more than 700 million tonnes of which about 70% is fly ash (Malhotra and Mehta, 2002). Huge quantum of coal ashes thus generated coupled with their very low specific gravity has made the ash handling and disposal problems very acute. A proper planning, sensible execution and good management of coal based thermal power generation projects will help not only in minimising the coal ash storage / disposal problems, but also in achieving many ‘positives’. This requires a better understanding of physical, chemical and engineering properties of coal ashes. This paper intends to critically evaluate these properties in general, the beneficial properties in particular, which have been hitherto considered as unwanted wastes and to suggest where exactly they can be effectively used in the field of geotechnical engineering. Most of the experimental data discussed in this paper are the outcome of the experimental investigation carried out at the Indian Institute of Science, Bangalore, on coal ashes procured from various thermal power plants distributed all over India. supported by Fly Ash utilization scheme of DST, Govt of India. Some of the data documented in the geotechnical engineering literature have also been made use of in the discussion.

12 ETGE 2012

1.1 Applications of coal ashes High Value utilizations: Mineral extraction; Ceramic industry; Floor and wall tiles; Acid refractory bricks; Fly ash distempers & paints; and Extraction of ceno spheres to name some of them. Medium value utilizations: Pozzolana cement; Cellular cement; Fly ash concrete; Fly as bricks & blocks; Prefabricated building blocks; Light weight aggregates; Grouting; Soil amendment agents/Fertilizers; and Soil stabilization, to name some of them. Low value but large scale utilizations: Mine filling; Back filling; Structural fills; Road construction; Mass concreting; and Embankment and dam construction. Considering the huge quantum of coal ashes being produced, the quantity of coal ashes being used in the non-geotechnical applications is negligible. Coal ashes appear to have attracted limited applications as construction materials except in some developed countries. Limited applications of coal ashes in the field of geotechnical engineering field hitherto can be attributed to lack of better understanding of the beneficial physical, chemical and engineering properties of coal ashes and the advantages they possess over fine-grained soils.

2. SOIL AND COAL ASHES Chemical compositions of coal ashes and soils are essentially similar (Table 1) except for the fact that the type of the trace elements present in coal ashes and soils can be quite different. However, coal ashes differ from soils on certain counts which favour their use as alternate / substitute materials to soil in the field of geotechnical engineering (Table 2).

Table 1. Range of chemical composition of Indian coal ashes and soils Compounds Fly ash: % Pond ash: % Bottom ash: % Soils: %

SiO2 38 -65 37 – 75 23 – 73 43 – 61 Al2O3 16 -44 11 – 54 13 – 27 12 – 39 TiO2 0.4 -1.8 0.2 - 1.4 0.2 - 1.8 0.2 – 2 Fe2O3 3 – 20 3 – 35 3 – 11 1 – 14 MnO 0 – 0.5 * - 0.6 * - 0.3 0 - 0.2 MgO 0.01 - 1.53 0.1 – 0.8 0.1 - 0.7 0.5 – 4 CaO 0.2 – 8 0.2 – 0.6 0.1 - 0.8 0 – 7 K2O 0.04 – 0.9 0.1 – 0.7 * - 0.56 0.3 – 2 Na2O 0.09 – 0.43 0.05 – 0.31 * - 0.3 0.2 – 3 L.O.I 0.2 - 3.4 0.1 – 7.91 0.61 - 12.8 5 – 17

LOI: Loss on Ignition at 9500C (includes H2O (+)) *: trace

Table 2. Differentiating factors between fine-grained soils and coal ashes Factors Natural fine-grained soils Coal ashes

• Nature of particles Mostly charged platelets No surface charges • Reactive silica content Almost nil Could be appreciable • Free lime content Nil Could be appreciable • Pozzolanic nature Mostly non-pozzolanic Could be appreciable

Majority of the fine-grained soils are physico-chemically active, which can be attributed to their unbalanced surface charges. Some of the undesirable properties of these soils such as high swell-shrink potential, high compressibility, low permeability, low strength and low CBR are essentially due to the surface charges of fine-grained soils. These surface charges

Coal Ashes in Geotechnical Engineering Practice 13

favour the development of diffuse double layer, which in turn can be taken to contribute to the water holding capacity of soils. As a consequence of this the volume stability of the soils gets adversely affected; effective voids volume gets reduced which is responsible for low permeability of such soil; actual void ratio becomes more than the theoretical which is responsible for higher compressibility; double layer repulsion increases which is responsible for the lower effective stress and hence, for lesser strength (Sridharan and Jayadeva, 1982; Sridharan and Venkatappa Rao, 1973; Sridharan and Venkatappa Rao, 1979; Sridharan and Prakash, 1999 Prakash and Sridharan, 2009).

3. BENEFICIAL PROPERTIES OF COAL ASHES 3.1 Specific Gravity While the values of specific gravity of soils vary over a narrow range of 2.55 – 2.8, those of coal ashes are found to vary over a wide range (i.e. 1.47 – 2.78). In addition, it has been found that the specific gravity of coal ashes is a function of their grain size (Pandian et al. 1998; Trivedi and Sud, 2004). Table 3 presents the representative values of specific gravity of coal ashes from different countries. Low values of specific gravity of coal ashes can be attributed to: spongy / porous nature of ash particles; presence of cenospheres; and unburned carbon content. Low to very low specific gravity of coal ashes makes them suitable for the use as the backfill materials in retaining wall construction, as construction fill materials on weak compressible soils, as fill materials for low-lying areas and as embankment materials. The advantages realised as a consequence of lower specific gravity of coal ashes in these applications are less lateral pressures on retaining structures

• less over burden pressures on foundation soils • reduction in the settlement of foundation soils • reduction in the tendency of weak sub soil to undergo failure • realisation of relatively steep side slopes of embankments

Table 3: Specific gravity of coal ashes from different countries

Type of coal ash Country Specific Gravity Reference Fly ash India

USA Canada Thailand

1.66 – 2.55 2.03 – 2.49 1.90 – 2.90 2.27 – 2.45

Author’s files Matin et. al. 1990 Indraratna et. al. 1991 Indraratna et. al. 1991

Pond Ash India UK Poland

1.64 – 2.66 2.10 – 2.24 1.90 – 2.31

Author’s files Skarzynska et. al. 1989 Skarzynska et. al. 1989

Bottom Ash India USA

1.47 – 2.19 2.28 – 2.78

Author’s files Seals et. al. 1972

3.2 Pozzolanic Reactivity Based on chemical composition, fly ashes have been classified into two groups namely class F and Class C fly ashes (ASTM C 618 – 94a). While class F fly ashes are pozzolanic, class C fly ashes have both pozzolanic and cementations properties. A pozzolanic reaction is one in which siliceous material reacts in the presence of moisture and calcium to form compounds exhibiting cementations properties. This property of coal ashes, particularly of fly ashes, makes them drastically different from fine-grained soils. The fly ashes are known for their

14 ETGE 2012

pozzolanic value as they are sources of reactive silica available in them in the amorphous form and / or as aluminate in the crystalline form. The pozzolanic reactivity or lime reactivity is normally expressed as the compressive strength of standard mortar cubes prepared using coal ashes and tested under specified conditions (IS: 1727, 1967). The lime reactivity of some typical Indian coal ashes are presented in Table 4. Fly ashes exhibit greater lime reactivity than the pond and bottom ashes due to their high reactive silica content. Table 4 also indicates that the lime reactivity reduces with aging.

Table 4. Lime reactivity values of typical Indian coal ashes*

Sl. No.

Source of coal ashes

Type of coal ashes

Lime reactivity: kPa

Fresh samples Aged samples

Vijayawada FA PA BA

3640 - -

2186 220 34

Ramagundam FA PA BA

4240 - -

4026 85 200

Farakka FA PA BA

5237 - -

1211 119 55

FA: Fly ash PA: Pond ash BA: Bottom ash * Data from author’s files The engineering performance of fly ashes gets improved with time, by virtue of the pozzolanic reactions. This property is responsible for them to exert lower lateral pressure on retaining structures, lower over burden pressures on foundation soils; to experience reduced secondary settlements and to have an increased shear strength and CBR with time. 3.3 Compaction Characteristics Compaction is an important process to which soil is subjected in the field to achieve the required dry unit weight at specified water content. The reference data for the field compaction is obtained through either standard Proctor or modified Proctor compaction tests in the laboratory. The test data is normally expressed through compaction curves (i.e. dry unit weight vs water content plot) along with zero air voids line. The range of specific gravity variation of coal ashes is more when compared with that of soils, in spite of identical chemical composition and grain size distribution. Hence, Sridharan et. al. (2001) felt that it would not be appropriate to compare the compaction characteristics of coal ashes with those of soils obtained through conventional compaction curves and that such a comparison would not be realistic (Fig. 1). They suggested the plotting of dry unit weights and corresponding water contents of coal ashes after normalising with a standard specific gravity. They suggested that 2.65 be taken as the standard value (Gstd) as it represented most of the soils. If γdn and wm are the dry unit weight and corresponding compaction water content of a coal ash of specific gravity Gm obtained from the compaction test, then the corresponding normalised dry unit weight and normalised water content can be calculated using equations 1 and 2.

γ=γ

m

stddmdn G

G (1)


0 20 40 60 805

10

15

20

25

Rihand (last field)

Dry

unit

weig

ht, k

N/m

3

0 20 40 60 805

10

15

20

25

Ramagundam

Water content, %0 20 40 60 80

5

10

15

20

25

Raebareli

0 20 40 60 805

10

15

20

25

Vijayawada

0 20 40 60 805

10

15

20

25

Z

X

BSES

0 20 40 60 805

10

15

20

25

Badarpur

0 20 40 60 805

10

15

20

25G

2.292.232.062.112.092.122.55

2.70

Source of Fly Ash

A-Z: Soils (Joslin, 1958)

Neyveli

0 20 40 60 805

10

15

20

25

0 20 40 60 805

10

15

20

25

0 20 40 60 805

10

15

20

25

0 20 40 60 805

10

15

20

25

0 20 40 60 805

10

15

20

25

0 20 40 60 805

10

15

20

25

0 20 40 60 805

10

15

20

25

0 20 40 60 805

10

15

20

25

0 20 40 60 805

10

15

20

25

0 20 40 60 805

10

15

20

25

TP

LH

DA

ZAV (G = 2.06)

ZAV (G = 2.70)

Figure 1. Compaction curves of typical fly ashes

(data source: Sridharan et. al. 2001)

0 20 40 60 805

10

15

20

25

Rihand (last field)

Norm

alise

d dr

y un

it we

ight

, kN/

m3

0 20 40 60 805

10

15

20

25

Ramagundam

Normalised Water content, %0 20 40 60 80

5

10

15

20

25

Raebareli

0 20 40 60 805

10

15

20

25

Vijayawada

0 20 40 60 805

10

15

20

25

Z

X

BSES

0 20 40 60 805

10

15

20

25

Badarpur

0 20 40 60 805

10

15

20

25G

2.292.232.062.112.092.122.55

2.70

Source of Fly Ash

A-Z: Soils (Joslin, 1958)

Neyveli

0 20 40 60 805

10

15

20

25

0 20 40 60 805

10

15

20

25

0 20 40 60 805

10

15

20

25

0 20 40 60 805

10

15

20

25

0 20 40 60 805

10

15

20

25

0 20 40 60 805

10

15

20

25

0 20 40 60 805

10

15

20

25

0 20 40 60 805

10

15

20

25

0 20 40 60 805

10

15

20

25

0 20 40 60 805

10

15

20

25

TP

LH

DA

ZAV (G = 2.65)

Figure 2. Normalised compaction curves for fly ashes

(data source: Sridharan et. al. 2001)

=

std

mmn G

Gww (2)

Fig. 2 represents typical compaction curves of Indian fly ashes in the normalised mode. Table 5 presents the compaction characteristics of coal ashes published in the literature. Fig. 2 suggests that the compaction curves and compacted dry unit weights are insensitive to the water content variation during compaction. These observations are of primary importance in that the field compaction does not require much of compaction control. This facilitates the

16 ETGE 2012

coal ashes to be effectively used in the construction of pavements and embankments. However, if the fly ash is of pozzolanic type (i.e. class C), then care should be exercised to avoid delay between mixing and compacting the fly ash in the field, as the delayed compaction results in lower dry unit weights and higher OMC (Sivapullaiah et. al. 1998).

Table 5. Compaction characteristics of coal ashes from literature

Type of coal ash Country γdmin kN/m3

Standard Proctor compaction Reference OMC: % γdmax kN/m3

Fly ash India Canada Thailand

8.0 – 15.5 – –

15.5 – 59.9 12.91 – 37.21 12.42 – 15.16

9.2 – 17.1 10.23 – 20.22 16.01 – 17.98

Author’s files Indraratna et. al. (1991) Indraratna et. al. (1991)

Pond ash India 7.8 – 15.8 14.6 – 36.8 12.2 – 17.1 Author’s files Bottom ash

India USA

7.6 – 12.6 8.89 – 13.29

21.3 – 58.1 14.2 – 23.2

7.5 – 13.7 12.98 – 17.10

Author’s files Seals et. al. (1972)

Note: The values of dry unit weight and OMC are normalised values. 3.4 Shear Strength Shear strength parameters depend upon the type of test, placement condition of the specimen and drainage conditions during testing. Table 6 gives the typical values of shear strength parameters of typical Indian coal ashes tested in shear box apparatus in the loose as well as compacted states. The study of shear strength behavior of coal ashes in the shear box apparatus reveals the following.

• Being cohesionless, non-plastic materials, coal ashes owe all their shear strength to frictional component except in the compacted, unsaturated state where apparent cohesion is also present which reduces to zero upon saturation.

• Coal ashes exhibit higher angle of shearing resistance, at both peak and residual stress levels, even in the soaked conditions.

• Coal ashes have angle of shearing resistance varying in the range 25o – 34o, even under loose conditions. The strength loss upon saturation is very small.

• In spite of their low unit weights, coal ashes exhibit high shear strengths when compared with natural soils.

Some typical results obtained from tri axial shear tests on Indian coal ashes are summerised in Table 6. The study of the shear strength behaviour of Indian coal ashes in triaxial testing apparatus both at the peak and at the residual test levels have indicated the following (Gray and Lin, 1972; Sridharan et al. 1998; Pandian et al. 2001b; Pandian et al. 2001c; Sridharan et al. 2002, Prakash and Sridharan, 2009).

• Variation of effective friction angle is negligibly small, irrespective of whether it is observed from consolidated undrained test or consolidated drained tests.

• Variation of effective friction angle of fly ash with initial dry density is not appreciable.

• For pozzolanic fly ashes, shear strength increases with curing period. • Peak and residual shear strength parameters are comparable. • Over consolidation increases the angle of shearing resistance appreciably.

High to very high shear strength parameters of coal ashes both at peak and residual stress levels, both in the loose condition and compacted / compacted – saturated condition favour their use in the field as all the problems concerned in the field with bearing capacity, slope stability of embankments, design of pavements and retaining structures are dependent on shear strength characteristics.


3.5 California Bearing Ratio (CBR) The CBR is an useful parameter in judging the suitability of the material for its intended use in the road construction and in the design of pavements. The CBR test can be carried out on

Table 6. Typical values of Shear Strength Parameters of Indian coal ashes at different states

(Sridharan et al. 1998)

Source of coal ash Type eloose

Loose state e at 95% γdmax

Compacted state

Compacted, saturated state

φ'dry, degrees

φ'res, degrees

φ'sat, degrees

c', kPa

φ', degrees

c', kPa

φ', degrees

Raebareli FA PA BA

2.36 1.38 2.60

31 32 32

29 30 31

30 30 31

0.66 0.96 1.51

23 16 17

34 31 32

00 00 00

33 30 31

Vijayawada FA PA BA

1.53 1.39 1.98

33 33 34

32 31 32

31 32 33

0.58 1.03 1.12

16 13 10

37 33 34

00 00 00

35 32 33

Badarpur FA PA BA

1.52 2.59 2.09

32 33 34

31 32 33

30 31 33

1.01 1.25 1.38

26 14 19

32 30 31

00 00 00

32 29 30

FA: Fly ash PA: Pond ash BA: Bottom ash Note: The values indicated are from box shear tests Table 7. Shear Strength Parameters of compacted and saturated coal ashes from triaxial shear

tests*

Placement condition

Type of coal ash

Consolidated drained tests Consolidated undrained tests

φcd: degrees ccd: kPa φcu:

degrees ccu: kPa φ': degrees c': kPa

Compacted to 95% γdmax. on dry side

FA PA BA

33-43 - -

00 - -

20-41 25-34 24-35

00 0-56 0-27

26-39 28-36 24-35

16-96 28-101 28-55

* Data from author’s files. compacted specimen either in the un soaked condition or after 96 hours of soaking. Both class F and class C fly ashes exhibit higher CBR in the un soaked condition. These higher CBR values are due to capillary forces, which exist in the partly saturated state. The CBR of class F fly ash tends to reduce drastically as the capillary forces reduce to zero on submergence. However, class C fly ashes retain very high CBR values even when tested after soaking, which can be attributed to pozzolanic reactions Normally, the design practice is to prefer soaked CBR values. However, it is justifiable to use soaked CBR values for those areas which are low-lying with poor drainage facilities, resulting in the submergence of roads. However, for areas that have good drainage facilities such that the roads will not get submerged even in the worst rains, it is justifiable to use the un soaked CBR in the design of pavements. The CBR of soils belonging to groups OH, CH, and MH have been observed to vary in the range 0-7% (Bowles, 1988). It is also observed that the CBR of coal ashes are much more than those of many fine-grained soils (Table 8). This characteristic makes them suitable for use as sub-base materials in the construction of pavements.

18 ETGE 2012

3.6 Compressibility and Consolidation Characteristics Compressibility characteristics namely compression index (Cc) and coefficient of volume change (mv) are important from the view point of calculation of settlement of structures.

Table 8. Values of CBR of compacted coal ashes and soils ( Prakash and Sridharan, 2009)

Sl. No.

Source of Material

Type of Material Testing condition

CBR: % Unsoaked

condition Soaked condition

Thailand FA (class C)

Compacted at OMC 325 280

Raichur FA Compacted at 0.95% γdmax, on the dry side

6.9 3.5

Vijayawada FA PA BA

Compacted at 0.95% γdmax, on the dry side

20.6 10.5 6.8

0.2 0.9 -

Raichur PA - do - 10.1 6.0

Badarpur PA BA - do - 11.1

11.3 4.4 8.5

Kahalgoan PA BA - do - 8.9

9.7 4.6 5.9

Davanagere Black cotton soil - do - 4.15 1.83

– Heavy clay Compacted at OMC 7.8 –

FA: Fly ash PA: Pond ash BA: Bottom ash Compressibility characteristics of fly ashes depend upon their initial dry unit weight, degree of saturation, self hardening characteristic, pozzolanic reactivity and mixing time (Gray and Lin, 1972; Yudhbir and Honjo, 1991). Their values along with the placement conditions are more meaningful while judging their suitability in the field than just their numerical values as depicted in Table 9. Fully saturated fly ashes are more compressible than the partly saturated fly ashes. Self hardening fly ashes compacted at OMC and saturated are less compressible than those compacted at OMC. If the fly ashes are of pozzolanic type, then the curing period also has appreciable influence on their compressibility (Yudhbir and Honjo, 1991).

Table 9. Cc as a function of placement void ratio

(Yudhbir and Honjo, 1991) Placement void ratio Placement condition Cc Comments

0.3 – 1.0 Conditioned compacted and conditioned grab placed

(0.1 ± 0.7) to (0.225 ± 0.15)

Very dense to medium dense

1.0 – 2.0 Hydraulically placed in lagoons

(0.225 ± 0.16) to (0.4 ± 0.21)

Loose to medium dense

2.0 – 3.0 Lagoon fly ashes to loose dumps

(0.4 ± 0.21) to (0.625 ± 0.26)

Very loose

Normally, 75% - 80% of total settlement of structures founded on fly ashes is due to primary consolidation, which depends upon their coefficient of consolidation (cv). The coefficient of consolidation of fly ashes is so high that it is extremely difficult to record time-compression readings in the laboratory consolidation testing to determine cv using curve fitting procedures.


In addition, it observed that the values of cv calculated from the curve fitting procedures from the laboratory very much underestimate the actual field behaviour. Hence, it is preferable to calculate the value of cv from the measured value of coefficient of permeability (k) and coefficient of volume change from equation 3. cv = k / [mvγw] (3) Table 10 lists the values of cv of coal ashes compacted at OMC from different countries. It has been observed that the values of cv of fly ashes from Hong Kong calculated using the field permeability with the help of eq. 3 are about 3 – 10 times higher than those listed in Table 10 (Yudhbir and Honjo, 1991). This lends support to the suggestion that the realistic values of cv can be obtained from eq. 3. In addition, the values of cv of fly ashes, which are silt sized particles, calculated from eq. 3 are in the range of values that corresponds to silts. Pond ashes and bottom ashes exhibit much higher cv values owing to their coarser size.

Table 10. cv of Coal ashes compacted at OMC

Country Type of coal ash cv: cm2/s Reference

*India FA 0.08-2 Kaniraj and Gaythri, 2004 FA 0.14-3.25 Author’s file

PA 0.96-10 Author’s file BA 1.43-10.15 Author’s file

Hong Kong FA 9.51 x 10-3

- 19.03 x10-3 Yudhbir and Hanjo, 1991

U.K FA 9.5 x10-4

- 6.34 x10-3 -do-

Thailand FA 3.2 x 10-4

-7.61 x 10-3 -do-

*cv calculated from measured coefficient of permeability using eq. 3. The higher values of cv of coal ashes signify an important fact that the primary consolidation of structures founded on coal ashes will be practically over during the period of construction itself. This feature makes the coal ashes superior for use as foundation base materials, as reclamation fills and as materials of construction for embankments and dams. 3.7 Permeability Coefficient of permeability of coal ashes depends upon their grain size distribution, testing conditions of coal ashes and the pozzolanic reactivity of coal ashes. Being coarser in size, bottom ashes are relatively more permeable than pond and fly ashes. Table 11 presents the values of k of coal ashes from different countries. The values of k of most of the fly ashes are in the range of k of silts. The permeability of coal ashes remains almost constant over a wide range of over burden pressure. These observations indicate that the coal ashes are normally freely draining materials, and are best suited for use as backfill materials behind the retaining structures, as sub-base materials in pavements and as embankment shell materials. The self cementing and pozzolanic fly ashes (i.e., class C type) exhibit lower permeability than class F fly ashes, and their permeability tends to reduce appreciably with time in the filed Such fly ashes can be more effectively used as liner materials in waste containment structures and as additives in the construction of effective seepage cutoffs like impervious blankets and cores in water retaining earth structures.

20 ETGE 2012

3.8 Swell and Shrink Potential Coal ashes exhibits high to very high volume stability (i.e., low swell and shrink potential), which can be attributed to their non-plastic nature and uniform gradation.

Table 11. Values of k for coal ashes from different countries

Country Type of Coal Ash

Testing condition of the coal ash

k: cm/s Reference

India FA PA BA

Compacted at γdmax and saturated

8 × 10-6 – 1.87 x 10-4 5 × 10-5 – 9.63 × 10-4

9.9 × 10-5 – 7.07 × 10-4

Author’s files

FA Compacted to 0.95 γd max and Saturated 1.4 ×10-5 – 4.23 × 10-4

Pandian and Balasubramonian, 1999

FA γd max 4.6 ×10-6 – 6 × 10-6 Kaniraj and Gayathri, 2004

Thailand FA (class C) Compacted at OMC < 10-7 Indrarathna et.al.,

1991 U K FA γd max

5 × 10-7 – 8 × 10-5 Gray and Lin, 1972 Japan FA Slurry

(ei = 0.85 to 1.02) 10-5 –10-4 Porbaha et.al., 2000

USA BA Relative density = 50 % 5 x 10-3 – 0.094 Seals et.al.,1972

Low carbon, high calcium FA (class-C) High carbon, low calcium FA (class-F)

γd max γd max

1 × 10-7 -2 × 10-7

6 × 10-5 – 2 × 10-6

Yudhbir and Honjo, 1991

FA BA

γd max γd max

1.8 ×10-5 -1.2×10-4 1.2 × 10-3 Martin et.al., 1990

Poland PA - 1.5× 10-5 – 5 × 10-5 Skarzynska et.al.,1989 Canada FA

BA In situ 10-7 – 10-4

3.4 x 10-3 – 4.8 x 10-3 Toth et al., 1988

Silt - - 1 × 10-7 – 1 x 10-3 - FA: Fly ash PA: Pond ash BA: Bottom ash

In the field of geotechnical engineering, the degree of expansively of soils can be judged based on Free Swell Ratio (FSR). It is defined as

FSR = k

d

VV

(4)

where Vd is the equilibrium sediment volume of 10 g of oven dried soil passing 425 µm sieve placed in a 100 ml jar containing distilled water with an initial volume of soil – water mixture equal to 100 ml after 24 hours of equilibration, and Vk is the equilibrium sediment volume of an identical soil sample in carbon tetra chloride or kerosene (Sridharan and Prakash, 2000b). Fine-grained soils can be classified as per the criteria given in Table 12. The Indian fly ashes, pond ashes and bottom ashes have been observed to have free swell ratios in the ranges 0.513 – 0.95, 0.647 – 1.1 and 0.8 – 1.16 respectively, indicating negligible degree of expansivity or swell potential. The shrinkability of soils is normally judged by their shrinkage limit. The shrinkage limit of soil is primarily controlled by the relative grain size packing of different sized particles


composing the soils (Sridharan and Prakash, 1998; Sridharan and Prakash, 2000a Prakash and Sridharan, 2009). Well-graded soils have lower shrinkage limits, and uniformly graded soils exhibit higher shrinkage limits. The non-plastic nature of coal ashes does not allow their shrinkage limit to be determined in the laboratory. However, it can be inferred that they exhibit high shrinkage limit owing to their uniform gradation.

Table 12. Soil classification based on FSR (Sridharan and Prakash 2000b)

Free swell ratio Clay type Soil expansivity ≤ 1.0 Non-swelling Negligible 1.0 – 1.5 Mixture of swelling and non-swelling Low 1.5 – 2.0 Swelling Moderate 2.0 – 4.0 Swelling High > 4.0 Swelling Very high

Low to very low swell and shrink potential of coal ashes can be taken the best advantage of in the construction of pavements, embankments, dams and as foundation base materials.

4. SOIL STABILISATION The poor gradation with silt / sand sized particles of fly ashes and their high to very high frictional strength even in the loose condition make them good mechanical admixtures in the filed of soil stabilisation. Addition of fly ash to cohesive soil will increase the strength of the resulting mix by virtue of the enhanced frictional strength and pozzolanic reactions. Fig. 3 shows the variation of CBR of a black cotton soil (wL = 56%; wP = 23%; wS = 10.3%; clay size fraction = 45.9%), which is an expansive soil, with the addition of different fly ash contents to the soil. Fig. 2 represents typical compaction curves of Indian fly ashes in the normalised mode. Table 5 represents the compaction characteristics of coal ashes represented in the literature. The study of Fig. 2 suggests that the compaction curves and compacted dry unit weights are insensitive to the water content variation during compaction. These observations are of primary importance in that the field compaction does not require much of compaction control. This facilitates the coal ashes to be effectively used in the construction of pavements and embankments. However, if the fly ash is of pozzolanic type (i.e. class C), then care should be exercised to avoid delay between mixing and compacting the fly ash in the field, as the delayed compaction results in lower dry unit weights and higher OMC (Sivapullaiah et. al. 1998). The addition of class F type of fly ash to the soil has resulted in a mix having more CBR than those of soil and fly ash alone. Both the curves, corresponding to soaked and unsoaked conditions of testing, exhibit two peaks. The first peak (represented by A and A') corresponds to soil stabilisation. This peak is a consequence of the following mechanisms. 1)Addition of fly ash provides coarser particles to improve the gradation of fine-grained soil. This will help in achieving better compacted density and hence, more strength. 2) Fly ash provides frictional component of shear strength to cohesive soil which has cohesive shear strength component already. This is responsible for the improved strength and CBR of the mix. These two mechanisms together help the stabilised cohesive soils in the field to exhibit a

22 ETGE 2012

better performance from the strength point of view. The second peak (represented by B and B') corresponds to fly ash stabilisation. This peak is due to the following mechanisms. The deficiency of finer particles in the fly ash is made up by the addition of fine-grained soil particles. This results is a better compacted density and hence, more strength Cohesive shear strength imparted by the addition of cohesive soil to fly ash has helped the resulting mix in achieving These two mechanisms together help the stabilised fly ash to exhibit higher strength and higher CBR. The CBR of class F fly ash can also be improved by the addition of coarser soil to it which results in a better grain size packing.

0 20 40 60 80 1000

2

4

6

8

10

12

B

B'

A

A'

Unsoaked condition Soaked condition

CBR:

%

Raichur fly ash content in the mix: %

Figure 3. Variation of CBR of BC soil – Raichur fly ash (class F fly ash) mixtures (data source: Pandian et al. 2001a)

The addition of class C type fly ash to a fine-grained clayey soil will continue to increase the CBR of the resulting mix with time due to the pozzolanic reactions (Fig. 4). In addition, the fly ashes when used as mechanical admixtures to stabilise expansive soil reduce the swell – shrink potential of expansive soils, thus providing them an improved volume stability. Apart from knowing these beneficial characteristics of coal ashes, one has to be aware of their certain undesirable properties also.

• Class F fly ashes are highly dispersive. With the result, they are easily erodible. • At very low compacted densities, they exhibit high collapse potential. • Their frost susceptibility is high.

However, these undesirable properties can be improved by treating them with chemical admixtures such as lime or cement or lime – gypsum and / or with mechanical admixtures such as soils.


.

0 20 40 60 80 1000

40

80

120

160

200

Unsoaked condition Soaked condition

CBR:

%

Neyveli fly ash content in the mix: %

Figure 4. Variation of CBR of soil – Neyveli fly ash (class C fly ash) mixtures (data source: Krishna, 2001)

5. CONCLUSIONS The common understanding among the people is that the coal ashes, which are by-products of thermal power generation industry, are waste materials which are harmful to the environment and to the people of the region as well. However, the study of the physical, chemical and engineering properties of coal ashes shows that the coal ashes are potential resourceful materials from the geotechnical engineering applications view point. The present paper has discussed many properties of coal ashes which can be used with the advantage in various geotechnical engineering applications. They are – low specific gravity, lower compressibility, higher rate of consolidation, higher frictional strength, higher CBR, negligible swell – shrink potential, water insensitiveness of compaction characteristics and pozzolanic reactivity. The beneficial properties of coal ashes discussed in this paper encourage their use as

• fill materials for low-lying areas • construction fill materials on weak compressible soils • The ever increasing scarcity for good materials in various geotechnical engineering

projects can also be overcome by the use of large scale use of coal ashes as • back fill materials in retaining structures • good foundation base materials • sub-base materials for pavements • construction of earth embankments and dams • mechanical admixtures in stabilising expansive and cohesive fine-grained soils.

ACKNOWLEDGEMENT The author thanks Prof. K Prakash, Professor of Civil Engineering, SJCE, Mysore for his help in preparing this paper and his contribution as co worker in many of his investigations. He is

24 ETGE 2012

grateful to the Indian Academy of Sciences for providing the Honorary Scientist position to him. The author wishes to thank Dr. Murali Krishna, IIT Guwahati for his help in formatting this paper. REFERENCES American Society for Testing Materials (1995),ASTM Designation C 618 – 94a, Standard

Specifications for coal ash and raw or calcined natural pozzolan for use as a mineral admixture in portland cement concrete, Annual book of ASTM standards, Vol. 104.02, ASTM, Philadelphia.

Bowles, J.E. (1988), Engineering properties of soils and their measurement, McGraw Hill Book Company, New York.

Gray, D.H. and Lin, Y.K. (1972), “Engineering properties of compacted fly ash”, J. Soil Mech. Found. Div. ASCE, Vol.98, No. SM 4, pp.361-380.

H.M.S.O. (1957), Soil mechanics for road engineers, Her Majesty’s Stationery Office, London.

Indian Standard Institution (1967), IS: 1727, Method of test for pozzolanic materials, BIS, New Delhi.

Kaniraj, S.R. and Gayathri, V. (2004), “Permeability and consolidation characteristics of compacted fly ash”, J. Energy Engineering, ASCE, Vol. 130, No. 1, pp. 18-43.

Krishna, K.C. (2001), CBR Behavoiur of Fly Ash - Soil - Cement Mixes, Ph.D thesis submitted to IISc, Bangalore, India.

Malhotra, V.M. and Mehta, P.K. (2002), High-Performance, High-Volume Fly Ash Concrete, Supplementary Cementing Materials for Sustainable Development Inc., Ottawa, Canada.

Martin, J.P., Collins, R.A., Browning, J.S. and Biehl, F.J. (1990), “Properties and use of fly ashes for embankments”, J. Energy Engineering, ASCE, Vol. 116, No. 2, pp. 71-86.

Pandian, N.S., and Balasubramonian, S. (1999), “Permeability and consolidation behaviour of fly ashes”, J. Testing and Evaluation, ASTM, Vol. 27, No. 5, pp. 337-342.

Pandian, N.S., Krishna, K.C. and Sridharan, A. (2001a), “California bearing ratio behaviour of soil / fly ash mixture”, J. Testing and Evaluation, ASTM, Vol. 28. No. 2, pp. 220-226.

Pandian, N.S., Rajasekhar, C. and Sridharan, A. (1998), “Studies of the specific gravity of some Indian coal ashes”, J. Testing and Evaluation, ASTM, Vol. 26. No. 3, pp. 177-186.

Pandian, N.S., Sridharan, A. and Chittibabu, G. (2001b), “Shear strength of coal ashes for geotechnical applications”, Proc. of Ind. Geotech. Conf. Indore, Vol. 1, pp. 466-469.

Pandian, N.S., Sridharan, A. and Chittibabu, G. (2001c), “Strength behaviour of compacted coal ashes for geotechnical applications”, Proc. of the International Symposium on Geotechnical and Environmental Challenges in Mountainous Terrain, Kathmandu, Nepal. pp. ?.

Porbaha, A., Pradhan, T.B.S. and Yamane, N. (2000), “Time effect on shear strength and permeability of fly ash”, J. Energy Engineering, ASCE, Vol. 126, No. 1, pp. 15-31.

Prakash, K. and Sridharan, A. (2006), “A geotechnical classification system for coal ashes”, Geotechnical Engineering, Proc. Inst. Civil Engg. (London), Vol. 159. No. GE2, pp. 91-98.

Prakash, K., and Sridharan, A. (2009), “Beneficial Properties of Coal Ashes and Effective Solid Waste Management”, Practice Periodical of Hazardous, Toxic, and Radioactive Waste Management, ASCE, Vol. 13, No. 4, pp. 239-248

Seals, R.K., Moulton, L.K. and Ruth, B.E. (1972), “Bottom ash: An engineering material”, J. Soil Mech. Found Div. ASCE, Vol. 98, No. 4, pp. 311-325.

Sivapulliaih, P.V., Prashanth, J.P. and Sridharan. A. (1998), “Effect of delay between mixing and compaction on strength and compaction properties of fly ash”, Geotech. Engg.


Bulletin, Vol. 7, No. 4, pp. 277-285. Skarzynska, K.M., Rainbow, A.K.M. and Zawiska, E. (1989), “Characteristics of ash in

storage ponds”, Proc. 12th Int. Conf. on S.M.&F. Engg., Rio de Janerio, Vol. 3, pp. 1915-1918.

Sridharan, A., Chittibabu, G. and. Pandian, N.S. (2002), “Strength behaviour of over consolidated fly ashes”, Proc. of Ind. Geotech. Conf. Allahabad, Vol. 1, pp. 3-6.

Sridharan, A., and. Jayadeva, M.S. (1982), “Double layer theory and compressibility of clays”, Geotechnique, Vol. 32, No. 2, pp. 133-144.

Sridharan, A., Pandian, N.S. and Srinivas, S. (2001d), “Compaction behaviour of Indian coal ashes”, Ground Improvement, Vol. 5, No.1, pp. 13-22.

Sridharan, A., Pandian, N.S. and Srinivasa Rao, P. (1998b), “Shear strength characteristics of some Indian fly ashes”, Ground Improvement, Vol. 2, No. 3, pp. 141-146.

Sridharan, A. and Prakash, K. (1998), “Mechanism controlling the shrinkage limit of soils”, Geotechnical Testing Journal, ASTM, Vol.21, No. 3, pp.240-250.

Sridharan, A. and Prakash, K. (1999), “Influence of clay mineralogy and pore medium chemistry on clay sediment formation”, Canadian Geotechnical Journal, Vol.36, pp.961-966.

Sridharan, A. and Prakash, K. (2000a), “Shrinkage limit of soil mixtures”, Geotechnical Testing Journal, ASTM, Vol. 23, No. 1, pp.3-8.

Sridharan, A. and Prakash, K. (2000b), “Classification procedures for expansive soils”, Geotechnical engineering, Proc. Inst. Civil Engg. (London), Vol.143, pp.235-240.

Sridharan, A. and Venkatappa Rao, G. (1973), “Mechanism controlling volume change of saturated clays and the role of the effective stress concept”, Geotechnique, Vol. 23, No. 3, pp. 359-382.

Sridharan, A. and Venkatappa Rao, G. (1979), “Shear strength behaviour of saturated clays and the role of the effective stress concept”, Geotechnique, Vol. 29, No. 2, pp. 177-193.

Toth, P.S., Chan, H.T. and Cragg, C.B. (1988), “Coal ash as structural fill with specific reference to Ontario experience”, Canadian Geotechnical Journal, Vol. 25, pp. 694-704.

Trivedi, A. and Sud, V.K. (2004), “Collapse behaviour of coal ash”, J. Geotech. and Geoenv. Engg. ASCE, Vol.130, No. 4, pp.403-415.

Yudhbir and Honjo, Y. (1991), “Applications of geotechnical engineering to environmental control”, Proc. 9th Asian Reg. Conf. on S.M.&F.E., Bangkok, Thailand, Vol. 2, pp. 431-469.

26 ETGE 2012


27

Ground Response and Support Measures for a Railway Tunnel in the

Himalayas

K. S. Rao Department of Civil Engineering, Indian Institute of Technology Delhi, Hauz Khas, New

Delhi, email: [email protected] ABSTRACT: The Pir Panjal tunnel linking between Banihal and Qazigund stations is the important tunnel in the railway line from Udhampur to Baramula in the Himalayas. The Pir Panjal ranges are having complex geological set up with major folds and faults. More than six major lithological units are traced along the 11 km length of the tunnel with very high overburden at many sections. The phenomenon of squeezing is studied using the limit equilibrium and FLAC methods for this tunnel. A detailed stress and displacement assessment has been attempted in this study, in order to stabilise the tunnel sections with suitable support measures. Keywords: ground response, support measures, railway tunnels 1. INTRODUCTION A large number of power and transport tunnel projects are being constructed in the tectonically active and young Himalayan Mountains. The main areas of concern regarding tunnel stability are the existence of weak, highly deformable and anisotropic rock mass and high degree of weathering and fracturing. Tunnel squeezing is common in the Himalayas in weak rock such as shale, slate, phyllite, schist and in weakness/fault zones and represents one of the major areas of concern regarding stability. Also in some areas due to very high overburden and brittle rock mass, explosive conditions develop resulting in rock bursts. Rock burst is the explosive failure in rock which occurs when very high stress concentrations are induced around underground openings. Though ravelling, swelling, running and flowing are occasional but rock squeezing is common in the Himalayas, leading to tunnel collapses. Several tunnels and bridges are being constructed by Ircon Int. Ltd and Konkan Rly Corp. Ltd for ambitious railway link between Katra to Qazigund in the Himalayas for the Northern Railways. The Pir Panjal rail tunnel is a part of the new railway line from Udhampur to Baramulla. The tunnel crosses the different formation of Pir Panjal range and runs around 11km length. In this study an attempt is made to evaluate ground response of Pir Panjal tunnel through limit equilibrium and Finite element approaches. Especially efforts were made to assess the squeezing and rock bursting conditions through out the length of the tunnel and stability measures are suggested for the affected sections. 1.1. Rockmass Response and Collapses Changes in stress around tunnel excavations can result in the behaviour of the rock mass which in turn may lead to damage, failure and consequent collapse of the rock mass. Sequence of rock mass behaviour leading to regional failure is explained schematically by Szwedzicki (2003) as shown in Fig.1. Accordingly, there will be several indicators and precursors which will lead to local damage and subsequently regional failure. An indicator is defined as a sign,

28 ETGE 2012

a state or a contributing factor that points out or suggest that the rock mass may be prone to damage or failure. In general potential failure is indicated by geotechnical and operational factors. A geotechnical precursor is a state or behaviour that suggests that the structure of the rock mass has been damaged prior to possible failure. Precursors, including results from instrumentation, warn of the development of excess ground deformations or high stress. Local damage is manifested by the following precursors e.g. spalling, squeezing, bursting, roof sagging, local falls, slabbing, joint dilation, creep, floor heaving, support damage etc.

Figure 1 .Sequence of Rockmass Behaviour 1.2 Squeezing Time dependent large displacement occurring around tunnels and other openings essentially associated with creeping is known as “squeezing rock conditions”. Several authors tried to explain the phenomena in the past (Dube et al. 1986, Ayden et al. 1993, Singh et al. 1999, Goel et al. 1995 and Panthi and Nilsen, 2007). Tunnels in weak rocks such as schists, shales, slates and phyllites when subjected to high ground stresses, experience squeezing conditions. This behaviour implies that yielding will occur around the tunnel resulting in convergence and face displacements. Singh et al. (1992) differentiated squeezing and non-squeezing overburden pressure and Q-ratting. Similarly several authors defined squeezing based on other criteria as well. Hoek and Marinos (2000) defined squeezing based on strain % (which is tunnel closure/tunnel diameter X 100) and ratio of rock mass strength and in situ stress. Based on this they proposed a classification for squeezing level as shown in Fig. 2. When σcm/po is low, the strain is very high (10%) indicating extreme squeezing conditions where as when the stress ratio is high (0 – 6), the strain % is < 1 % implying few support problems. The above classification is used in this study to define the squeezing problems. 2. THE PIR PANJAL TUNNEL The proposed Pir Panjal tunnel is part of the railway line from Udhampur to Baramula in the Himalayas. The tunnel across Pir Panjal range is located between the future railway stations Banihal in the south and Qazigund in the north. The total length of the horse shoe shape with flat floor tunnel will be 11 km (10960m) length with 8.0m height and 8.94m width. It is

Ground Response and Support Measures for a Railway Tunnel in the Himalayas 29

completely straight and runs almost parallel to North-South direction. The overburden at both portals is about 10m above tunnel crown, while the maximum overburden is approximately 1150m. About 4 km of the tunnel length has an overburden of more than 500m and about 650m of the length has more than 1000m. The excavation method is NATM with drill-blast. The tunnel layout is shown in Fig. 3.

Figure 2. Classification of Squeezing

Figure 3. 2D View of Pir Panjal Tunnel 2.1 Geological Setup The tunnel alignment traverses through steeply slopping highly undulating hill slopes of the Pir Panjal range which is part of the young lower Himalayas. Formation levels at tunnel portals are at elevation 1713.63 (South portal) and 1956. 70 (North portal). The highly folded and faulted mountain ranges have a strike of bedding is NW-SE. Distinct folding is visible in the central regions. Bedding of the southern slopes dip with 60 – 90⁰ towards NE while on northern slopes dip with 36 – 45 towards SW. Contact between rock units are often faulted. The lithological units are Zewan beds, Gangamopteris bed, Panjal traps, conglomerate bed, agglomerate slates, fenestella shale and syringothyeis limestone. The main rock units are limestones, quartzites, shales, sandstone, conglomerate and fluvoglacial materials. Table 1 details the anticipated rock units at different chainage of the tunnel. The table also presents maximum and minimum overburden at different sections of the tunnel.

30 ETGE 2012

Table 1. Anticipated Rock Types Over the Length

2.2 Geotechnical Parameters Extensive geotechnical investigations were carried out through drilling number of boreholes as well as several shafts and drifts. Because of high overburden, there was a limitation of drilling depth. However, based on surface mapping, drilling and mapping in the drifts and adits, the rock type classes were determined. Extensive laboratory tests were carried out for obtaining bulk density, cohesion, c, friction φ, uniaxial compressive strength, σc , tensile strength, σt, Modulus, Et and Poisson’s ratios for all rock varieties. Rock mass properties and joint parameters were also established through relevant field and laboratory tests. Adopted geotechnical parameters for the ground response analysis are presented in Table 2. 3. GROUND RESPONSE ANALYSIS: ANALYTICAL APPROACHES Predictions of stresses and displacements around a circular opening in rock mass at great depth are an important problem in geotechnical, petroleum and mining engineering. The main analytical approaches adopted are:

i. Limit Equilibrium Method ii. Numerical Method


The closed form solutions are based on simplified assumptions e.g. shape of the opening is regular (mostly circular, elliptical, or spherical), the media is homogeneous and isotropic. They are easy and provide insight into how the mechanical variables influence the deformation behaviour (Hoek and Brown, 1994). To identify the magnitude of stresses and deformations in Pir Panjal tunnel, calculations were carried out based on closed form solution for circular shape of equivalent opening in elasto-plastic medium with primary stress field of Ko=1.

Table 2. Geotechnical Design Parameters

Numerical methods include such techniques as finite element, finite difference and boundary element. Depending upon geological media two approaches to numerical modelling is identified. A continuum approach treats the rock mass as continuum intersected by a number of discontinuities, while a discontinuum approach views the rock mass as an assemblage of independent blocks or particles (Goodman and John, 1977). Further, continuum models are of two types: differential and integral. Differential models characterise the entire region of interest and include the finite difference and the finite element methods. Where as integral or boundary element models feature discretisation only along interior or exterior boundaries. Ground response in the study was obtained using the powerful FLAC method. The Fast Langrangian Analysis of Continua (FLAC) is a two dimensional explicit finite difference program. In order to setup a model to run a simulation with FLAC, in the fundamental components of the problem shall be specified: a finite difference grid, constitutive behaviour and material properties and boundary and initial conditions. The general solution procedure as indicated in FLAC manual version 5.0 is adopted for the study. The Mohr- Coulomb model which is convenient is used in the study. The descretised model of the Pir Panjal tunnel obtained by FLAC is shown in Fig. 4. Because of axisymmetric half of the tunnel is descretised into 15876 square and rectangular zones. The boundary conditions are applied in terms of both stresses and displacements. The bottom boundary is fixed in Y-direction where as the left vertical boundary is restrained in X- direction. A vertical stress Syy and the horizontal stress, Sxx are applied on top boundary and vertical right side boundary. Stresses and deformations before and after support system are calculated using both the methods.

32 ETGE 2012

Figure 4. Descretized Model of Tunnel

4. RESULTS AND DISCUSSION Analysis based on closed form solutions for the six varieties of rock masses available along the Pir Panjal tunnel has been carried out and stresses and displacement are obtained. Variation of radial and tangential stresses with distance from the centre of the tunnel and also variation of displacements (radial) with distance are plotted for all rock types. Typical such variations for shale are shown in Fig. 5(a) and (b). As per the figures, it is clear that the σθ = 19.56 MPa occurs at 10.5 m from the centre of tunnel, where as radial stress is zero at the boundary. Maximum deformation of 59.6 m was observed at the tunnel boundary and also radius of plastic zone is 10.5 m. Results for the rocks at different sections are summarised in Table. 3. For prediction of squeezing behaviour, strains were calculated from the displacements and presented in Table 3 for all six types of rocks. As shown in table agglomeratic shale and shale show 1.03 and 1.33% of strain respectively indicating moderate squeezing where as all other rock types show no squeezing.

Table 3. Comparison of Analysis


Stresses and displacements are also obtained from FLAC and corresponding strains were obtained for all rocks. The results are shown in Fig. 5 (a), (b), (c) and (d) for shale and similar graphs were plotted for other rocks as well. The results are compared in Table 4 along with the results obtained by closed form solutions. It is clear that the deformation values obtained from FLAC are higher than the closed form solutions. The strains for both rocks fall under moderate squeezing category. Rock mass response behaviour from closed form solutions and FLAC obtained for all rock masses. FLAC software is used to stabilise the tunnel at all sections with shotcrete lining and rock bolts of appropriate input parameters. After installation of support shortcrete lining is checked in bending and direct stresses and rock bolts are checked in tensile stresses.

Figure 4. (a) Stress Variation (b) Deformation in Shale

Table 4. Details of Stresses

SHALE 1100

0

5

10

15

20

25

0 10 20 30 40 50

Distance from centre (m)

Str

ess

(MP

a)

Radial stress

Tangentialt

SHALE 1100

0

10

20

30

40

50

60

70

0 10 20 30 40 50

Distance from centre (m)

Ra

dia

l dis

pla

cem

en

t (m

m)

Radial displacement

(a)

(b)

34 ETGE 2012

Bending movements, axial force and structural displacements in supports for all rock types are plotted and typical results obtained for shale are presented in Fig. 6 (a) and (b). Comparison of results without and with support is given in Table 4 and 5 for all sections.

Table 5. Comparison of Results

Figure 5. (a) Maximum Principal Stress (b) Minimum Principal Stress (c) Y Displacement (d) Plasticity Indicator Contours for Shale

(b) (a)

(c) (d)


Table 5 shows the shotcrete lining of 300 mm thickness is not safe in bending stresses in shale at section 1100/22. Therefore, the thickness of shotcrete needs to increase. After installation of supports all sections except shale are stabilised. Shale is suffering from moderate squeezing. For such condition, forepoling and advance face stabilisation are required. Yielding support system may be required in extreme cases to prevent squeezing conditions.

Figure 6. (a) Axial Force and (b) Displacement in Shale

5. CONCLUSIONS The Pir Panjal tunnel in the Himalayas traverses through much diversified geology experiencing high ground stresses. Time dependent behaviour of tunnels in squeezing rock is investigated using Closed form and FLAC methods. Stresses and displacements are obtained before and after the installation of support measures. Nominal shortcrete and rock bolting proves ineffective to control squeezing in shales, hence advance face stabilization approaches have been suggested in this work. REFERENCES Dube, A.K., Singh, B. and Singh, B. (1986), Study of squeezing pressure phenomenon in

tunnel-1, Tunneling and underground space technology, Vol. 1, No. 1, pp. 35-39.

FLAC (Version 5.00)

LEGEND

19-May-08 18:01 step 1124 6.000E+01 <x< 9.000E+01 6.000E+01 <y< 9.000E+01

Grid plot

0 5E 0

Cable Plot# 2 (Cable) -1.651E+02# 3 (Cable) -8.960E+01# 4 (Cable) -4.662E+01# 5 (Cable) -3.084E+02# 6 (Cable) -3.898E+01# 7 (Cable) -8.220E+01# 8 (Cable) -9.909E+01# 9 (Cable) -9.116E+01#10 (Cable) -8.020E+01#11 (Cable) -8.463E+01#12 (Cable) -9.072E+01#13 (Cable) -1.006E+02#14 (Cable) -8.736E+01#15 (Cable) -4.364E+01

6.250

6.750

7.250

7.750

8.250

8.750

(*10 1̂)

6.250 6.750 7.250 7.750 8.250 8.750(*10 1̂)

JOB TITLE : Pir Panjal Tunnel in Shale VI H=1100 m

FLAC (Version 5.00)

LEGEND

19-May-08 13:09 step 927 6.000E+01 <x< 9.000E+01 6.000E+01 <y< 9.000E+01

Grid plot

0 5E 0

Structural DisplacementMax Value = 5.091E-03

6.250

6.750

7.250

7.750

8.250

8.750

(*10 1̂)

6.250 6.750 7.250 7.750 8.250 8.750(*10 1̂)

JOB TITLE : Support Installation

(a)

(b)

36 ETGE 2012

FLAC (2005) (version 5.0) Tutorial manual Goel R.K., Jethwa, J.L., and Paithankar A.G. (1995), Tunnelling in the Himalayas- problems

and solution, Tunnels and Tunnelling, Vol. 27 (5), pp.58-59. Goodman, R.E. and John (1977), Finite element analysis of discontinuous rocks, Numerical

methods in Geotech Engg., Mcgrow-Hill, Newyork. Hoek, E. and Brown, E.T. (1980) Underground excavations in rock, pp. 1-15. Hoek, E. and Marioons, P. (2000), Predicting tunnel squeezing problems in weak

heterogeneous rock masses, Tunnels and tunnelling international part 1, pp. 1-20. Jethwa J.L., Singh B. and Singh B. (1984), Estimation of ultimate rock pressure for tunnel

lining under squeezing rock conditions – A new approach. Design and performance of underground excavations, ISRM symposium, Cambridge, E.T. Brown and J.A. Judson eds., pp. 231-238.

Singh, M., Singh B. and Singh, J. (2006), Critical strain and squeezing of rock mass in tunnels. Tunnelling and underground space technology, Vol. 22(3), pp. 343-350.

Szwedzicki, T. (2003), Rock mass behaviour prior to failure. International journal of rock mechanics and mining sciences, Vol. 40, pp. 573-584.


37

Treatment of Foundations and Geological Faults of Almatti Dam on

Krishna River: A Case Study

Mahavir Bidasaria Managing Director, Ferro Concrete Const. (India) Pvt. Ltd., Indore.

email : [email protected] & [email protected] ABSTRACT: Effective Foundation treatment is an important component of construction of dams and other hydraulic structures. Curtain Grouting is an important part of Foundation treatment of Dam and other hydraulic structures. The purpose of Curtain Grouting is to form a zone of low permeability upto a designed depth on the upstream of the dam. This grout curtain along with downstream drainage system controls the uplift pressure and piping which are potential hazards. The present case study deals with providing grout curtain to Almatti Masonry Dam from Block No. 1 to 52. Besides curtain grouting this case study also deals with the treatment of geological fault (unconformity zone) of thickness upto 6 m, which was existing in the foundation from Block No. 45 to 52 below the joint of base granite rock and overlying quartzite foundation. The single line grout curtain of permeability less than 3 Lugeon has been effectively formed below the foundation, despite very poor strata having maximum pre-grout permeability of 90 Lugeon. Keywords: foundation treatment, geological faults, dam, grouting,

1. INTRODUCTION Geology of dam foundation poses many surprises and have to be dealt with, according to the prevailing options. Curtain Grouting is an important component of Foundation Treatment for all the high Dams and water retaining structures. The purpose of curtain grouting is to form a zone of low permeability upto a designed depth below a specified portion of the upstream of the dam. Under reservoir condition, this zone forms a total or partial barrier to seepage flow below the dam. This barrier i.e. grout curtain along with downstream drainage system controls the uplift pressure and piping which are potential hazard for a water retaining structure. The present case study deals with “Curtain Grouting and drilling drainage holes for Almatti Dam”. Besides dealing with curtain grouting through drainage gallery from Block No. 1 to 52 of Almatti Dam one of the main objective of this work was also to treat the weak zone (unconfirmly zone) which was existing in the foundation from Block No. 45 to 52, between the joint of base granite rock and overlying quartzitic formations. 2. SALIENT FEATURES OF ALMATTI DAM At present Almatti Dam is practically completed on Krishna river at Almatti Village in Bijapur Distt. of Karnataka, India. It is the main reservoir for Krishna valley projects of Karnataka. Height of this dam is 40 m from lowest bed level of river. The length of masonry portion of dam is around 1162 meters and earthen dam on left flank is around 402 m. It has been fitted with 26 Nos. of Crest Gates on spillway of size 15 m x 15.24 m.

38 ETGE 2012

3. DESIGN OF GROUT CURTAIN AND DRAINAGE HOLES Design of grout curtain is based on factors like geology of dam foundation hydraulic head, the potential of leakage and piping and off course it is based on the observation and performance of grout curtain of other dams. Parameters like geometric layout i.e. spacing of holes depth of grout curtain, limit of reduction of permeability below dam foundation are selected keeping in mind the factors as explained above. In present case these parameters were decided by Dam Safety review panel (DSRP) of this project as below: Layout: Single line grout Curtain with primary holes spaced at 3m c/c. and secondary or tertiary holes till desired permeability is obtained. Depth of Grout Curtain - 0.5 H. H = Hydraulic Head from foundation. Permeability Limit of Grout Curtain - < 3 lugeon. Drainage holes:

(a) Depth of Drainage hole - 0.75 of depth of Grout Curtain.

(b) Spacing of d/Holes - 3 m c/c. The location of grout curtain and drainage holes is shown in Fig. 1 4. EXECUTION OF GROUT CURTAIN AND DRAINAGE HOLES Single row grout earlier has been provided from Block No. 1 to 52, through drainage gallery. Primary holes were drilled and grouted at 3 m c/c and after completion of primary holes, secondary holes are drilled and grouted at 1.5 m c/c. In few blocks heavy artisan conditions were observed, even tertiary holes were required to be done to bring down the permeability below 3 lugeon. Every 10th hole has been drilled first by NX diamond core, to get the geotechnical details of the foundations beneath the dam, before curtain grouting work is taken up in hand block by block. Length of foundation gallery of this dam from block 1 to 52 is 1111.50 m. To understand the complex geology of dam foundations and curtain grouting treatment of this dam through drainage gallery it is found necessary to deal it in three sections blockwise i.e. Section I form Block 1 to 30, Section II from Block 31 to 44 and Section III from Block 45 to 52. these sections are formed looking to the diverse behaviour of dam foundations during curtain grouting treatment. Geological Section of Dam Foundations is shown in Fig. 3. Details are discussed below section wise. Section I –from block no. 1 to 30 1.0 Geology of dam foundation : 1.1 In all about 43 holes were drilled by core drilling (NX size) practically at 15m c/c. from block No. 1 to 30. The bores have established fresh and hard migmatite granite gneiss associated with later intrusions of pneumatic bands, micro granite and dolerite dykes. In most of the bore holes Core recovery is very good and is ranging from 80% to 100%.

Treatment of Foundations and Geological Faults of Almatti Dam 39

1.2 The Cores reveals that the granite formations are intersected by various sets of joints of which the sub-horizontal joints, varying between 70 deg to vertical are very much predominant and joint planes are observed invariably in cores even in deeper horizons.

1.3 Rock was very hard and abrasive in nature. Pre-grout permeability is normal and

ranging from 3 Lu to 24 Lu.

40 ETGE 2012

2.0 Drilling and curtain grouting (1 to 30) : 2.1 Drilling of Primary holes at 3 m c/c were taken up first. These holes were grouted in

stage of 10m. The holes were washed thoroughly after drilling of a stage/holes is completed, by alternate jets of Air and water.

2.2 In all the holes pre-grout permeability tests were taken. It was observed that on an

average pre-grout permeability was 12 Lu., in most of the blocks. The maximum permeability of 29 Lu was found to be in Block No. 15 in first stage.

2.2.1 It has been observed that Grout intake on an average, came to 35 kg/m of drilling.

However Maximum grout intake was 1065 kg. in 1st stage of one Hole in block No.2.

After completion of Primary Curtain holes, Secondary holes were drilled at 1.5m c/c. The procedure of drilling, washing, water intake tests, grouting etc., were same as done in primary holes.

2.2.2 In all the Secondary holes, pre-grout permeability were taken. It was observed that on

an average pre-grout permeability was 7 Lu. The maximum pre-grout permeability in secondary hole was found as 22 Lu., in 1st stage of one hole in block No.11

2.2.3 It was observed that grout intake on an average came to 20 kg/m of drilling. However

maximum intake was 210 kg/m in stage III of holes in block No. 29. 2.2.4 The grout pressure in various stages are taken as below,

Ist stage 10m - 3.5 kg/cm2 at Gauge. IInd stage 10 to 20m - 6.5 kg/cm2 at Gauge. IIIrd stage 20 to 30m - 10.0 kg/cm2 at Gauge and onward stage

2.2.5 The consistency of grout has been taken from 1:10 to 1:1 Cement to water by weight 2.3 After completion of curtain grouting primary and secondary holes in Block No. 1 to 30,

one test hole was taken in each block and stage wise post grout hole drilled in between the primary and secondary holes where grout intake is maximum. The result reveals that permeability values have come within permissible limit (less than 3 to 5 Lu), in all the blocks from 1 to 30

2.4 After completion of curtain grouting in Block No. 1 to 30, in all respect, and ensuring

that all the blocks, post grout permeability has reduced less than 3-5 Lu, DRAINAGE HOLES were taken up downstream of grout curtain at 3m c/c. After drilling these drainage holes, they were washed thoroughly with the jet of air and water, and capped.

Section II – from block no. 31 to 44 1.0 Geology of dam foundation

In all about 19 holes were drilled by Core drilling practically at 15 m c/c from Block No. 31 to 44. The cores reveals that the granite formations are intersected by various sets of joints of which the sub-horizontal joints varying between 70 deg to vertical are very much pre-dominant and joints are open. It has practically the same geology as Block no. 1 to 30 but it looks has more criss cross joints and fault zones which are highly water bearing. While going through bore logs it is observed that many diamond drill cores had poor recovery, mainly in Block No. 32, 39 and 40.

2.0 Drilling and curtain grouting (artisan conditions in holes) :


Drilling and Curtain grouting was performed in single line as done from Block 1 to 30. But, while drilling primary holes in these blocks, Artisan conditions were observed. It was observed that while drilling primary holes in block No. 31, 32 and 33 during March to May 99, when water level in reservoir was minimum on upstream of dam, these holes at the time of drilling showed severe Artisan conditions and jet from drill holes practically touching the roof of the gallery. Practically in all the blocks from Block No. 31 to 44, there were artisan conditions.

After grouting of primary and secondary holes completed, one test hole was taken in each block, and the following observations were made,

i. In block no. 32, 35 and 36, test hole after secondary hole grouting showed the artisan conditions.

ii. In test hole in Block No. 32, after secondary hole grouting Lugeon value found to be 11.36 in second stage.

The observations made above were discussed with the Dam Safety Review Committee meeting and after review of the grouting results, it was recommended to take up tertiary holes in these blocks, in between the primary and secondary holes.

Accordingly, tertiary holes were drilled and grouted in these blocks, and one test hole was taken in each block. The results have revealed that the post grout permeability in these blocks is within the limits, and the artisan conditions have completely disappeared.

Section III – from block no. 45 to 52 1.0 Geology of dam foundations :

In all about 12 holes were drilled by core drilling (NX size) at 15 m c/c. These Bore Holes reveal as below:- i. In this reach from Block No. 45 to 52, the bores have shown existence of

Quartzite overlaying the basement granite formations see Fig. 2

ii. A weak zone (unconformity zone) lying between Quartzite and the basement granite on the right bank has been established. A typical type of yellow soil weathered as well as fused formations have been observed practically in every bore hole. Thickness of this unconformity zone is varying from 0.5 m to 6 m thickness.

iii. The bores have picked up deeply weathered/disintegrated zones in the Quartzites

in the bottom lifts.

iv. In most of the Diamond drill holes it is reported that the drill rods were dropping suddenly and it has recorded poor core recovery and highly fractured Quartzites.

2.0 Drilling and curtain grouting:

During drilling of primary holes in these blocks, again heavy artisan conditions were observed.

Primary and secondary drilling and grouting of Block No. 51 and 52 were completed.

In these blocks Quartzite formations is overlying basement granite formations. Between these two formations, there is a weak zone (unconformity zone) of 0.5 m to 6 m depth where core recover is very poor (10 to 25%) and permeability is also very high (around 20 to 90 Lu).

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Having a complex geology of these reaches, it is necessary to analyze Primary, Secondary and Tertiary drilling and grouting separately as below:-

2.1 Primary hole curtain grouting (block no. 45 to 52)

The work of curtain grouting was taken up in these blocks, in the similar method as in the earlier blocks. The primary holes are taken up 3.0 m c/c initially. However, in most of the primary holes artisan conditions are met with and jets are rising upto the roof of the gallery. In primary holes, the maximum permeability to the tune of 90 Lu was found in stage VII in a hole in a Block No. 51, and the maximum grout intake was 3200 Kg in this stage.

2.2 Secondary hole curtain grouting (block no.45 to 52)

After completion of primary holes drilling and grouting, secondary holes drilling and Grouting were taken up at 1.5 m c/c i.e. between the already done primary holes.

The maximum pre-grout permeability of 78 Lugeon was observed in a hole in Block No. 48. However pre-grout average permeability of 22 Lugeon was recorded.

2.3 Tertiary hole curtain grouting (block no. 45 to 52)

The Dam Safety Review Committee during their visits reviewed the observations made during the drilling and grouting of primary and secondary holes in these blocks, and the geology of the strata beneath the foundation in these blocks. They recommended to take up tertiary holes in between the primary and secondary holes in these blocks, and to take up grout holes up to basement granite.

Tertiary holes drilling and grouting had been started and completed from Block No. 45 to 52.

It was found that maximum pre-grout permeability of 19 Lugeon was recorded in a Block No. 48. Further pre-grout average permeability as recorded is 10 Lugeon and average grout intake as 54 Kg/mt.

2.4 Post grout test holes :

After completion of tertiary holes, NX size test holes have been done one in each Block from 45 to 52 and water intake test in holes were conducted. It was observed that Lugeon value has come down well within limit i.e. maximum 3.00 Lugeon and minimum 1.62 Lu.

2.5 Further drainage holes in these blocks had been taken up and completed. 5. CURTAIN GROUTING FROM THE UPSTREAM HEEL OF DAM Looking to the poor Geotechnical condition, it was recommended to provide another line of grout curtains from the upstream heel of Dam from Block No. 45 to 49, to strengthen the foundations in view of the relatively thick unconformity zone observed. Accordingly, the drilling and grouting was taken up, and the test hole results have revealed permissible limits. 6. GROUTING IN TRANSVERSE GALLERY IN BLOCK NO. 45 The transverse gallery has been provided in Block No. 45 for treating the geological faults existing across the Dam. The DSRP recommended to drill grout holes along the fault line in


the transverse gallery and grout them to effectively treat the major fault. As per the recommendations, two lines of grout holes along the fault zone were provided in the transverse gallery. 7. CONCLUSIONS The single line curtain has been completed in all the blocks, and it was observed that the grout curtain of permeability less than 3 Lugeon has been effectively formed below the foundations, despite of very poor strata (unconformity zone) having maximum pre-grout permeability of 90 Lu. This single line grout curtain has been formed from drainage gallery, in all the blocks from 1 to 52 and the same was further strengthened by providing second line of grout curtain extended from upstream of dam in block Nos. 45 to 49 having very poor geology. This grout curtain is further supported by providing drainage holes downstream of Grout curtain at 3m c/c.

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45

Recent Experiences of Ground Stabilization Techniques

Satyendra Mittal

Department of Civil Engineering, Indian Institute of Technology Roorkee, Roorkee email: [email protected]

ABSTRACT: Ground Improvement is the need of hour due to limited space available to meet the demand of large population. Therefore now a days, construction of undersigned parkings, metro rails, highways etc. have been quite common. In such situations, only innovative ground improvement techniques can work. In the older period also people used to adopt ground improvement techniques in their own way. Soil Nailing, reinforced gabion walls have been successfully used by author at several sites. The same are described in the present paper. Keywords: ground stabilization, ground improvement, soil nailing, landfills

1. INTRODUCTION With the rapid industrialization and increase in population, a lot of construction activities are taking place. Hence, an engineer is compelled to accept all types of land whether soft or good. Gone are the days when people used to reject any site on account of being water logged, full of compressible soil or otherwise. Now the technology is available to improve all types of soils. The concept of earth reinforcement is not new, the basic principles are demonstrated in nature by animals and birds and the action of tree roots. The fundamentals of techniques are also described in Bible covering the reinforcement of clay or bricks with marsh plant or straw for the construction of dwellings. Constructions using these techniques were existed in 5th & 4th millennia B.C. Gautam Budha through his favorite disciples constructed houses for people who were homeless by use of torn off and unused clothes (cut into small pieces), mixed into clay. The clothes’ pieces evidently increased the tensile strength of soil and thus the homes built by Budha survived for a long time. The earliest remaining examples of soil reinforcement are Great Wall of China, Agar – Quf Ziggmat, situated 5 km north of Baghdad. This was constructed with clay bricks varying in thickness between 130-400 mm, reinforced with woven mats of reed laid horizontally on a layer of sand and gravel at vertical spacing varying between 0.5 and 2.0m. Reeds were also used to form a rope (100mm Φ) which pass the structure and acted as reinforcement. This structure is 45m tall (originally it is believed to have been over 80m high) and is 3000 years old. The Great Wall of China is another example of reinforced soil where mixture of clay & gravel was used with tamarisk (bushes grown along sea side) branches. The Romans are also known to have used this technique. Pasley (1822), a colonel in British army constructed retaining walls by reinforcing the backfill with brushwood, wooden plank. Reinforced earth finds application in construction of dams, railway lines, river training works, dykes etc. In 1930s, French scientist constructed a ladder wall wherein retaining wall consisted of mass of granular filling unified by a row of tie members each having a small end anchor, together with a thin cladding membrane, as shown below (Fig. 1):

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Figure 1. Ladder wall

2. GROUND IMPROVEMENT The ground improvement is the technique to improve soil characteristics to improve its engineering performance as per the project requirement by altering its natural state, instead of having to alter the design in response to the ground natural limitations. The ground improvement techniques are used to: -

· Reduce the settlement of structures. · Improve the shear strength of soil and thus increasing the bearing capacity of soil. · Increase the factor of safety against possible failure of embankment. · Reduce the shrinkage and swelling of soils. · Increase the liquefaction resistance.

3. HISTORICAL EVIDENCES OF GROUND IMPROVEMENT With the invention of wheel in the land of Sumer about 35000 years ago, the mobility of man was greatly enhanced leading to the establishment of settlement and the inter connection of societies. In those days builders often faced great difficulties in the road construction over the marshy and unsuitable land without the facility of modern equipments. Soil reinforcement technique was applied for the construction of roads over soft soils, using wooden logs by the builders (2400 BC). The same technique was also later adopted by Romans (Fig.2). The Sumerians also used the concept of consolidation in Mesopotamia. At the Agarquf Ziggmat (1400 BC) in Iraq, reed were bound together to form cables about 100mm in diameter that acted as horizontal drains for dissipation of pore water pressure build in the soil mass as earth structure increased in height. Romans used timber piles as reinforcement to provide the load – support, to allow the construction of roadways over very soft soils, in Italy. The concept used by the Romans when crossing marshes along their 800 km network of roads is shown in Fig. 3.

Recent Experiences of Ground Stabilization Techniques 47

Figure 2. Technique adopted by Romans

The early accounts of densification go back to ancient India, where elephant were used for compaction of soils by treading. The use of animals such as cows, oxen and calves for soil densification was specified in the old Hindu set of building rules written in Manasara Shilpashastra, (written in Sanskrit). The first use of dynamic compaction goes back to the days of Roman Empire. In chapter III of Book III of his De Architectura, on Roman architecture, Vitruvius writes, if, the soil is not firm, it should be made firm by dropping a weight. Ground densification methods were described by the Chinese in the Dung Dynasty code of 1103 AD. Some of these methods are illustrated in Fig. 4.

Figure 3. Compaction Technique Chemical stabilization was effectively used by the Romans for the construction of their famous roads. They improved their road building capabilities by introducing natural pozolanic soils, lime and even furnace slag as construction material. Many examples were found in ancient road by the Romans in Italy and England (Goodchild, 1956).

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Figure 4. Ground Densification (1103 AD.) 4. VARIOUS METHODS OF GROUND IMPROVEMENT If the virgin soil is soft and replacement of existing soil is not feasible, in that case it is recommended to treat that soil at site itself. if, the surface soil is granular, the ground improvement usually consists of densification. If the soil is cohesive, the ground improvement techniques applied may include consolidation, weight reduction or reinforcement using vertical reinforcing element. If the weak soil is very close to the surface, densification with heavy vibratory roller, chemical stabilization with admixture or reinforcement with geosynthetics can be used. Ground improvement methods can be divided into eight main categories (Table -1).

Table 1. Methods of Ground Improvement

Sl. No. Methodology Nature of Work

1 Removal of soft soils

Removal of unsuitable soil

2 Preloading of soils

Pre load the existing soil prior to construction

3 Stage(s) construction Construction of embankment in parts/stages

4 Densification Static compaction, vibroflotation, dynamic compaction, stone columns, blasting and compaction grouting

5 Consolidation Preloading, wick drain, vacuum consolidation 6

Weight reduction Use of Fly ash as light material, wood chips, slag, geofoam, tyre chips

7 Reinforcement Use of geosynthetics, Soil nailing, micro-piles,

8 Chemical treatment Jet or fracture grouting, admixtures mixing, deep mixing.

9 Biotechnical stabilization Brush layering, brush matting, composite system


5. SOIL NAILING Soil nailing is a technique where either natural soil or existing fill material is reinforced by the insertion of tension elements called soil nails. Soil nails are made of metallic or polymeric material and may be installed into a pre-drilled hole and then grouted, drilled and grouted simultaneously, or inserted using a displacement technique. The nails are installed horizontally or at a slight downward inclination to the horizontal. A hard, flexible or soft facing may be used at the surface of the slope. The main components of a typical soil nail are shown in Figure 5. Some common applications of soil nailing are given as below and also shown in Figures 6 & 7.

Figure 5. A general view of soil nail

Figure 6. Application of nailing in house/road application

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Figure 7. Nailing in transmission tower Construction

5.1 Advantages and Limitations of soil nailing The advantages and limitations of soil nailing are summarized in Table 2. 5.2 Installation Methods Soil nails are commonly installed by one of the following methods:

· bored then grouted · self-drilled · driven or rotated.

The other steps are as follows:

1. Install the nails 2. Fix the head plates and nuts to the nails. 3. Place the welded wire mesh. 4. Apply the sprayed-concrete facing. 5. Excavate the next step. 6. Install the drainage (for example, pipes and geotextile bands) as appropriate, as well as

structural items. 6. INTERNAL AND EXTERNAL STABILITY CHECKS As with all earth retention systems and slopes, the stability of the nailed cut also needs to be assessed and confirmed. In general, a limit equilibrium analysis is used because this is relatively simple and gives satisfactory results. Numerical modeling methods such as the finite element method may also be used, but these are complex. They also require high-quality site investigation data coupled with careful calibration of the constitutive model to give meaningful and reliable results. The following external failure modes should be considered in the analysis of soil-nailed walls or slopes:

· stability against overturning · sliding failure (shear at base) – for steep slopes and walls only · bearing failure (basal heave) – also only for steep slopes and walls


Table 2. Advantages and Limitations of soil nailing Advantages Limitations Construction equipment Drilling rigs for nail installation and guns for applications of sprayed concrete (where required) are relatively small, mobile and quiet. This is an advantage in urban areas where noise, vibration or access may pose problems. It is also of benefit for infrastructure earthworks slopes with limited access. Cost Soil nailing can result in significant savings compared to conventional techniques. Construction flexibility Soil nailing can proceed rapidly and the excavation can be shaped easily. It is a flexible technique, rapidly accommodating variations in soil conditions and work programmes as excavation progresses. Performance Measurements indicate that the overall movements required to mobilize the reinforcement forces are usually small and similar to other earth-retention systems. Environmental and aesthetic considerations Soil nailing can provide an aesthetically pleasing solution.

Temporary stand-up time For new steep cuts, soil nail construction requires the formation of cuts generally 1-2 m high in the soil. These then have to stand unsupported for at least a few hours before placement of sprayed concrete and nailing. The soil therefore has to have a degree of natural cohesion or cementing. Excavation below the water table Generally unsuitable for soil nailing. Conventional embedded retaining wall techniques are better suited. Dewatered face This is desirable for soil nailing. If the groundwater percolates through the face the unreinforced soil will slump locally on initial excavation, making it impossible to establish a satisfactory sprayed concrete face. Excavation in Soft Clay Unsuited to soil nailing because the low frictional resistance of the soft clay would require a very high density of soil nail reinforcement of considerable length to ensure adequate levels of stability. Vertical or steep excavations in high-plasticity clays adjacent to movement-sensitive buildings or services The risks of using soil nailing in these cases should be carefully considered. Soil nailing may offer less security than conventional embedded retaining wall techniques in these cases. Nails may be of a length that necessitates additional land-take This may constrain future development of land above the nails. Seasonal shrink/swell movement in high-plasticity clays There is a lack of appropriate design techniques that deal directly with, and lead to proper control of, these movements – and therefore a lack of understanding of how to avoid associated degradation and onset of progressive failure.

7. SIMPLE DESIGN OF NAILED WALL Schlosser (1982) found that the active failure zone for nailed slopes was similar to, but larger than, that of a reinforced soil wall. In both cases, the active failure zone is smaller than the standard Coulomb active wedge assumed with the other retaining structures. It has been suggested by him that this difference in behaviour is attributable to the inclination of the soil nails. The reduction in the lateral earth pressure, as a result of the inclined face, should enable the length of the nails to be slightly less than that for the reinforcement to a vertically faced

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reinforced soil wall. It is therefore often possible to use a soil nail length of about 60-70% of the height of the sloping face. In view of the limited number of structures built to date, it is suggested that a relatively high value of 2.5 is adopted as the minimum acceptable overall adherence safety factor. The overall adherence design of a reinforced soil wall involves calculation of the bond developed on the length of reinforcement beyond the potential failure plane. For the case of either no surcharge, or a uniform surcharge, the potential failure plane can be considered to be inclined at 90 – ½(b+f) to the vertical, where b is the inclination of the face to the vertical. At first sight, it might seem logical to take the friction developed on a soil nail as simply the product of the vertical stress by the active length by the soil nail perimeter by tan m, where m is the angle of friction between the soil and the soil nail. However, this does not take account of the fact that the horizontal stress acting on the sides of the soil nail will be less than the vertical stress acting on the top and bottom of the nail. Referring to Fig. 8, the nail friction is given as:

FN = La tan mgh¢d[2+(p-2)Ko] (1) Where, La is the active length of nail beyond the failure plane, d is the diameter of the soil nail, h’ is the depth of the soil above the active adherence length of the nail, Ko is the coefficient of earth pressure at rest. The tension developed in each soil nail can be assumed to be the lateral earth pressure acting over the area of the slope face that is supported by the soil nail. Schlosser (1982) observed that, in the case of both soil nailed slopes and reinforced soil walls, the reinforcement restrained lateral movements, particularly at the top of the structure. In this region, the lateral earth pressure coefficient is therefore close to Ko. However, these structures tend to have a fairly uniform reinforcement distribution, making the lowermost reinforcement the most critical level in the tension design calculations. For this location, the earth pressure coefficient can safely be assumed to be Ka. Therefore the tension in the soil nail (T) is given by

T = KaghmSV (2) Where

hm is the depth of the lowest soil nail V is the vertical spacing between the soil nails S is the horizontal spacing between the soil nails.

For cohesive soils, laboratory tests should ideally be carried out to determine the surface adhesion, using a specially modified shear box. In the case of geotextile ties with a ribbed surface, the adhesion coefficient can often be assumed to be equal to the lesser of 50kN/m2 or the soil cohesion value. During construction, this assumption should be checked by conducting pull-out tests. If these indicate that the adhesion between the nail and the soil has been over-estimated, then the length of the nail should be increased to compensate for this deficiency. The spacing of soil nails obtained with the Gassler and Gudehus (1983) design charts is generally fairly close to that obtained with the adopted reinforced soil wall procedure. A third alternative design approach described by Schlosser (1982) is a computer program known as `TALREN’ which was developed in France. This considers the equilibrium of a circular slip surface through the nailed soil slope, using a method of slices rather similar to the


Fellenius and Bishop methods used with unreinforced slopes. As the contribution of the soil nail’s bending resistance, shear strength and tensile strength is considered in the computer program, it is more suited to grouted steel soil nails than to geotextile soil nails. A software MSNAILS has been developed by Mittal (2006) by use of which nailed cut can be designed within a short span of time. Gosavi et al. (2006) have also developed the design charts considering the log-spiral rupture surface. By using the design charts, the nailed wall can be designed easily. In the absence of these charts, following method can be used. 7.1 Design Example

It is proposed to form an 8m high soil nailed slope excavated in a cohesionless soil with f = 36o and g = 18 kN/m3. The face of the slope is to be inclined at 10o to the vertical and the soil nails are to be inclined 10o below the horizontal. The nail ties have a short-term ultimate strength of 50 kN, and a safety factor of 3.0 on this strength will ensure a design life of 120 years. The nail diameter is 15mm and their surface friction coefficient is 80% of that of the soil. If the layers of soil nails are to be spaced at 1m vertical intervals, starting at a depth of 0.5m below the upper soil surface, determine the required horizontal spacing and length of the soil nails.

Solution Using equation below, Ka can be determined.

2

5.0)(sinsin5.1)(sin

)(sin

úúú

û

ù

êêê

ë

é

+

-=

bfb

fbaK (3)

2

5.0)80(sin36sin5.1)80(sin

)3680(sin

úúú

û

ù

êêê

ë

é

+

-= ooo

oo

= 0.191 Using Ko » 1 – sin f » 0.4122 Neglecting the anchoring force developed on the small driving heads at the end of the soil nail, and taking the bond as equal to the shaft friction, gives FN = La tan mgh¢d[2+(p-2)Ko] (In this equation, tan m can be replaced by m’ also = 0.8 tan d as it is given that friction coeff. Is 80% of that of soil). As tan m = 0.8 tan 36o FN = La 0.8 tan 36o x 18h’ x 0.015 [2+(p - 2)0.4122] = 0.388 Lah’ Inclination of the potential failure plane to the vertical = 90 - ½ (d + f) = 32o Referring to Figure 5, the length of soil nail within the potential failure wedge (L’) is given by

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Figure 8. A general cross section of nailed slope and its front view

Figure 9. Typical view of a nail

41.0)8(10cos

22tan)( ho

ohHL -=-=¢

in D ABC, Cos 10o = (H-h)/AC in D ACM, L’ = (H-h) tan 22/Cos 10. Nail depth at M, h’ = h + L’ tan 10o (here tan 10o can be replaced by Sin 10o also as for small angles, sin q = tan q.

Trying a soil nail length of 70% H = 5.6m, cautiously ignoring the increase in nail depth beyond point M, and tabulating the overall adherence calculations gives the following (Table. 3).


Table 3. Design calculations Level Depth h L’

(8-h) 0.41 H’ = h + L’ tan 10o

La=5.6-L’ Friction = 0.388 Lah’

1 0.5 3.08 1.04 2.52 1.02 2 1.5 2.67 1.97 2.93 2.24 3 2.5 2.26 2.90 3.34 3.76 4 3.5 1.85 3.83 3.75 5.57 5 4.5 1.44 4.75 4.16 7.67 6 5.5 1.03 5.68 4.57 10.07 7 6.5 0.62 6.61 4.98 12.77 8 7.5 0.21 7.54 5.39 15.77 å58.87 kN

Permissible tension force = 50/3 = 16.67kN (Here 3 = F.O.S.). Therefore, as none of the friction forces (given in last column of above table) developed exceed this value, the total frictional bond for a single vertical row of soil nails = 58.87kN. Lateral earth pressure per m = ½KagH2 = ½(0.1981) x 18 x (8)2 = 114.1kN Trying a horizontal soil nail spacing of 0.2m gives the lateral earth pressure per vertical row of soil nails = 114.1 x 0.2 = 22.82kN Hence the overall adherence safety factor = 58.87/22.82 = 2.6 (adequate) Using equation for determination of Tension for the tension design of the soil nails gives T = KaghmSV = 0.1981 x18x7.5x0.2x0.5 = 2.67 kN Therefore the tensile safety factor = 50/2.67 =18.7 (adequate). Thus nailed wall can be designed like this design example. Author has successfully tried soil nailing at various sites in Hero cycles industry Ludhiana, Tunneling work in Delhi, Allahabad, Amritsar, Landslide controls in Nainital, Shimla, Mussorie, Dehradun etc.

8. STABILITY STUDY OF ASH POND DYKE WITH SPATIAL CONSIDERATIONS (MITTAL & WAYAL, 2012)

Millions of tons of fly ash/pond ash is generated from thermal power stations, disposal of which is a serious problem. A little quantity of this fly ash is being used in cement industry, mining, road construction but even then the substantial quantity of ash remains unutilized. On every thermal power station the dykes are created to store the pond ash. When the first proposed height of dyke (mother dyke) is filled up then the subsequent raising is done over it in form of trapezoidal shape dyke. A stage comes when the 1st raising is also filled up though it comes after nearly 2 years or so, depending- upon the height of raising and the area available at the site.

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When the 1st raising is filled up, it might affect the stability of mother dyke because of unprecedented loads, seepage, vehicular moments etc. The current practice is to make the raising over previous raising (called 1st raising), laid on the mother dyke. The 2nd raising is also done in the form of trapezoidal dykes, same as that in case of 1st raising. The 2nd raising also has substantial weight besides the seepage pressure, vehicular loads (for compaction etc.) which might further jeopardize the stability of not only the 1st raised dyke but also the mother dyke. Thermal power station authorities have reported to author the cases of destabilization of mother dykes at some sites due to 2nd raising. The methodology is basically based on the design suggested by Prof. M.R. Madhav for one of the thermal power stations in the country. Mittal (2012) conducted further analysis in this direction and gave detailed design to another thermal power station in country. The latter has incidentally approved the design suggested by Mittal (2012). The stability analysis was conducted using Geostudio software -SLOPE/ W [5]. In addition, the DC-Gabion software [6] was used to analyze the stability of Gabion wall constructed to retain the additional ash. These computer models were used to assess the stability of the dyke in terms of factor of safety (FOS). The analysis was performed for the determination of a slope stability factor of safety with computational modeling. This study has been done in following headings: Stability of Mother Dyke alone: For this study the model of following dimension was prepared in Slope/W software. The foundation of 10 m thickness is provided below mother dyke. For controlling internal erosion properly designed drainage and filter system is provided in the dyke. Rock toe has also been provided. Phreatic line of water is also drawn as ash slurry contains some water .The properties of the ash material, rock material and foundation soil are given in next chapter.

· Stability of Mother dyke with 1st raised ash pond dyke: Again model was prepared with the raised ash dyke of height of 6m and the stability analysis was done to obtain critical slip circle and FOS.

o Stability study of Mother Dyke with 2nd raised ash pond dyke o Stability measures of Mother dyke with 1st raising. o Stability measures of Mother dyke with 2nd raising.

· Creation of extra space for storage of ash within arena of mother dyke: This was

done by providing Gabion wall around dyke. The height of gabion wall was same as height of mother dyke i.e. 8.7m. And the stability of Gabion wall was then done in DC-Gabion software.

· Creation of extra space for storage of ash within arena of Mother Dyke and 1st

raised dyke: Gabion wall of height 13.7 m taken for study which was analyzed in DC-Gabion software.

· Creation of extra space for storage of ash within arena of Mother Dyke and 1st

and 2nd raised dyke: In this study the Gabion wall of height 12m was designed at heel of first Raised dyke and this was then analyzed in DC-Gabion software. The details of design are available elsewhere (Mittal, 2012). Two given vide Figs. 10 and 11 for the benefit of readers.


Figure 10. Model of mother dyke with 1st raised dyke and with gabion walls (13.7 m and 6m

high)

Figure 11. Gabion wall of height 13.7 m

8.1 Retaining Wall with Reinforced Gabion Boxes It is an innovative design based on reinforced earth wall concept. Only difference is that in this type of wall, instead of precast concrete blocks, the wall is constructed of gabion boxes. Each box is made of 1m x 1m cross-section. The length of box may be 1m to 4m. The stability analysis of such wall may be done either manually or any commercial software. The length of reinforcement (behind the wall) in the backfill is normally 0.7 times the full height of wall. The only exception in this type of wall is that GI wire of gabion box also goes into backfill for generally 0.35 times height of wall. The GI wire of box is PVC coated to protect the wire against corrosion. Author has designed many such walls in Chandigarh, Mussorie, Cochin etc (Fig.12).

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Figure 12. (a) A cross section of gabion reinforced earth wall

9. CONCLUSIONS All the methods described in this paper are actual case studies which have been implemented at various sites by author. The gabion walls in ash dykes provide a passive support to ash dyke and thus allow further raising of dykes. The case study discussed here is for a thermal power project where millions of tonnes of flyash is produced every year and govt. has made very strict laws against purchase of new grounds for storage of flyash. The gabion walls discussed in paper are also provided due to availability of very limited space at site. Both the methods proved to be much more economical than purchase of new grounds.


Figure 12. (b) Specifications and details of gabion wall

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REFERENCES Murray, R.T. (1993), “The development of specifications for soil nailing”, Research Report

380, TRL Crowthorne. Gassler, G. & Gudehus, G (1983), “Soil nailing – statistical design”, Proc. 10th Eur. Conf. Soil

Mech. & Found. Engg., Helsinki, Vol.2, pp 491-494. Schlosser, F. (1982), “Behaviour and Design of Soil Nailing, Proc. Symp. On recent

developments in ground improvement techniques, Bangkok, AA Balkema, Rotterdam. Gosavi, M., Saran, S. and Mittal, Satyendra (2006), “Software Development for design of

nailed open Cuts”, Proc. Of Indo-Australian Conf. On Information technology in Civil Engg., IIT Roorkee.

John, NWM, “Geotextiles”, Blackie. Soil nailing best practice guidance – CIRIA Publishers. Saran, S., Mittal, Satyendra and Gosavi, Meenal (2005), “Pseudo Static Analysis of Nailed

Vertical Excavations in Sands”, Indian Geotechnical Journal, 35(4), 401-417. Mittal, Satyendra (2006), “Soil Nailing Application in erosion control – an experimental

study”, Journal of Geotech & Geological Engg, (Springer), pp 675-688. Mittal, Satyendra & Wayal, Vaishali (2012), “Stability study of ash pond dyke with spatial considerations”

Journal of Structural Engg. – Accepted for publication.


61

Performance Based Earthquake Resistant Design of Geotechnical

Structures – A New Trend

S. K. Prasad and P. Nanjundaswamy Department of Civil Engineering, S. J. College of Engineering, Mysore

email: [email protected] ABSTRACT: Performance based design of structures is gaining considerable importance in earthquake engineering in recent times. It is being established that earthquake resistant design based on displacement criteria is more acceptable than strength based design. There are considerable efforts by geotechnical engineers to adopt this philosophy. Considering the complexity and uncertainty of soil behavior, and randomness of earthquakes, it is even more worthwhile to adopt displacement based approach for geotechnical structures subjected to seismic action. This paper emphasis the need for performance based design of geotechnical structures subjected to earthquake loading. In this connection, concept of performance based design, Pushover analysis, geotechnical considerations for performance based design are explained. Further, a work carried out at S. J. College of Engineering, Mysore to develop and validate the analytical model for evaluating the coupled permanent displacement in sliding and tilting of quay wall is presented. More precise determination of permanent deformations of structures and identifying the degree of damage to the system at different stages of deformation are important components of performance based design. Keywords: performance based design, earthquake, geotechnical structures, quay walls

1. INTRODUCTION Concept of performance based design Current method of seismic design of structures favors performance based approach either for the design of new structure or for rehabilitation of existing structure. Typically, a performance objective is defined when a set of structural and non-structural performance levels, representing damage, loss and repair or rehabilitation cost are related with the different intensities of seismic input. Naturally, type of construction, quality of materials used, efficiency of construction workers and engineers, amount of money spent and importance of structure among many influence the performance level. The present earthquake resistant design methods emphasize ductile design of structures wherein the structure is expected to withstand a smaller earthquake elastically and an unexpected earthquake by undergoing inelastic loading cycles sustaining its integrity. The performance of a structure is typically assessed on the basis of maximum deformation and cumulative inelastic energy absorbed during the earthquake. Reports from the observations during past earthquakes, model testing on shaking table and analytical modeling of behavior of structure during earthquakes suggest that most structures designed according to the present codal practice will sustain residual or permanent displacement under a design earthquake even if they perform exactly as expected. Residual or permanent displacement can result in partial or total loss of a structure such that it becomes unfit to withstand the design static load in future. It may be unsafe to the occupants or may not withstand the subsequent earthquakes under the new at rest position. Further, the cost of repair or replacement of structural element can be different for varied residual

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displacements. All these factors are not properly brought out in the current force based design approach. However, residual displacement is unavoidable in a structure that experiences inelastic deformation under severe seismic shaking. Performance based design is an emerging methodology, which was developed after the lessons learned from earthquakes in the 1990s. The goal is to overcome the limitations present in conventional seismic design. Conventional seismic design is based on providing capacity to resist a design seismic force, but it does not provide information on the performance of structure when the limit of the force-balance is exceeded. If the limit equilibrium is not exceeded for relatively high intensity ground motions associated with a rare seismic event, the construction/retrofitting cost will most likely be too high. On the other hand, if force balance design is based on a more frequent seismic event, then it is difficult to estimate the seismic performance of the structure when subjected to ground motions that are greater than those used in design. The static design of structures includes limit states of collapse and serviceability. Limit state of collapse is force based design wherein the stresses are determined under the design load in the entire body of the structures and the design ensures that the stresses are within the permissible limits. Limit state of serviceability is the displacement based design wherein the strains and displacements are maintained well within the permissible limits. In both the limits the quantities are well within the elastic limits of the material of structure. The dynamic design under earthquake loading has a different concept. A structure is usually designed under two levels of earthquake force. Level I design involves the design basis earthquake. It is the ground motion which the structures are likely to experience at least once in the life time of structure having a return period of 250 years and 10 % probability of occurrence. Level II design involves maximum credible earthquake having a return period of 500 years and probability of occurrence of 2%. A performance based seismic design procedure requires a quantification of performance, which is based on one or multiple structural response indices. Traditionally, ductility, energy dissipation or their combination is identified as the parameter that evaluates the performance level of structural element. In performance based design procedure, appropriate levels of design earthquake motions must be defined and corresponding acceptable levels of structural damage must be clearly identified. The acceptable level of damage is specified according to the specific needs of the users or owners of the facilities and may be defined on the basis of the acceptable level of structural and operational damage. The structural damage category is directly related to the amount of work needed to restore the full functional capacity of the structure and is often referred to as the direct loss due to earthquakes. The operational damage category is related to the length of time and cost associated with the restoration of full or partial serviceability. Economic losses associated with the loss of serviceability are often referred to as indirect losses. In addition to the fundamental functions of servicing sea transport, the functions of port structures may include protection of human life and property, functioning as an emergency base for transportation, and as protection from spilling hazardous materials. If applicable, the effects on these issues should be considered in defining the acceptable level of damage. Once the design earthquake levels and the acceptable damage levels are properly identified, the required performance of a structure can be defined by appropriate performance grades. In performance based design, a structure is designed to meet these performance grades.

Performance Based Earthquake Resistant Design of Geotechnical Structures – A New Trend 63

The principal steps taken in performance based design are: A. Selection of a performance grade: Either based on the acceptable level of damage

consistent with the needs of the users or owners of the facility or on the basis of importance of the structure. Degrees of importance are defined in most seismic codes and standards.

B. Defining damage criteria: Specifying the level of acceptable damage in engineering parameters such as displacements, limit stress states, or ductility factors. These damage criteria related to engineering parameters are obtained from the experience of the past earthquakes or laboratory tests and subsequent analytical simulations.

C. Evaluation of seismic performance of a structure: Evaluation is typically done by comparing the response parameters from a seismic analysis of the structure with the damage criteria. This allows direct comparison between the required performance grade and the seismic response of the structure in question. If the results of the analysis do not meet the damage criteria, the proposed design or existing structure should be modified.

The above mentioned design steps are shown in the flow chart in Fig. 1.

Figure 1. Flowchart for seismic performance based design of structures under seismic conditions

Yes

No

Acceptable Damage Serviceable, repairable, near collapse, collapse

Earthquake Level Level 1 or Level 2

Performance Grade

Damage Criteria

Seismic Analysis

Response Parameters

Damage criteria satisfied

End of Design/Evaluation

Modification

A

B

C

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2. PUSHOVER ANALYSIS Pushover analysis is a simplified nonlinear analysis on structural members subjected to incremental lateral loads until failure. The central focus of this analysis is the generation of the pushover curve or capacity curve which brings out the relationship between base shear generated (lateral force applied to the structure) and corresponding roof displacement in lateral direction until collapse. This capacity curve is representation of the structures ability to resist the seismic demand. To generate the capacity curve, the structure is pushed in a representative lateral load pattern which is applied monotonically while the gravity loads are in place. Any type of representative lateral load pattern can be defined but the load pattern similar to first mode shape amplitude of the structure is the most commonly used to determine the capacity. For a given structure and ground motion, the displacement demand is an estimate of maximum expected response of building during ground motion. Once capacity curve and demand displacement are defined, a performance point can be determined. A performance check verifies that structural and non-structural components are not damaged beyond the acceptable limits of the performance objective for the forces and displacements implied by the displacement demand. Fig. 2 presents the most essential part of identifying and classifying the different earthquake performance levels for any structure. From fully operational to collapse of structure during earthquake, it is possible to divide in to different states. The performance level also depends on the intensity of earthquake considered. Under smaller earthquakes, the structure may perform very well and may remain within elastic limits. However, under a bigger earthquake, the same building may be subjected to higher levels of damage and life safety, near collapse or total collapse states. Hence performance levels and degree of importance of structure should be kept in mind in performance based design. For this purpose, a more realistic structural model is chosen. Under a given force, the deformation is estimated as shown in Fig. 3. At various stages of displacement, different performance levels such as elastic, fully operational immediate occupancy, life safety, collapse prevention and total collapse are identified. The limits for each state are defined. Then, depending on the capacity of a particular structure, under the given force, a particular state is reached. This approach helps in identifying the performance of either an existing structure or a structure newly designed.

Figure 2. Combination of performance levels and earthquake hazards (Vision 2000)


Figure 3. Performance check through Pushover analysis

Fig. 4 presents a typical pushover curve popularly used in assessing the capacity of structural frame in resisting earthquake force. The frame to be analysed is idealized, a typical model to define the behavior of structure is identified, and earthquake force is simplified as a lateral force at the base of the structure which is increased from zero till the failure of structure is identified. As shown in the figure a graph of base shear carried by the structure to the corresponding roof displacement is plotted and the different states are identified. Depending on the level of earthquake, a demand curve is obtained and the intersection point of capacity curve and demand curve gives the performance point. This point signifies the state of the structure under a design earthquake.

Roof Displacement

Figure 4. A typical pushover capacity curve with different states

3. PERFORMANCE BASED DESIGN OF GEOTECHNICAL STRUCTURES The geotechnical structures comprising of earthen slopes, earth embankments, retaining walls and quay walls, tunnels, foundations possess more uncertainties and complexities as their behavior is governed by their individual performance, performance of soil around and the interaction between the two. The presence of water can alter the soil behavior enormously. Complex earthquake loading can enhance the difficulty level in understanding the performance. Further, the force based approach can help in determination of factor of safety against failure without considering the actual displacements. In geotechnical structures displacement or strain level attained plays a major role in estimating the performance. The following are the major points to be considered in the performance based design of geotechnical structures. 3.1 Choice of Performance Grade It is of utmost importance to choose a particular type of performance grade. Performance Grade depends on many factors such as the importance of a particular structure, frequency of occurrence of earthquake, the magnitude of earthquake, the impact of the event socially and

A

Base Shear

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economically and the economic conditions of the region in addition to technical expertise of engineers and labour force. Essentially, it is the owners or the end users who should decide on the desired level of performance. 3.2 Definition of damage criteria This is the most important parameter in the performance based design. It is important to define the criteria of damage as insignificant, moderate, considerable, sufficiently large, enormous, collapse, catastrophic based on the extent of damage, amount of loss suffered, difficulty level for repair and loss of life. The amount of time required for restoration, repair or rehabilitation is the other factor which defines the damage criteria. In most of geotechnical problems, the level of strain experienced by the ground plays an important role in assessing damage. Hence, residual displacement/tilt experienced by the structure can be a parameter which defines the damage criteria. Engineers, owners and end users will own the responsibility in deciding the damage criteria. 3.3 Evaluation of Seismic Performance This task lies in the hands of engineers. The performance of the concerned structure under earthquake force should be assessed. For this purpose, the engineer can estimate the mechanism of performance, mode of failure and evaluate the maximum strain and corresponding displacement. Depending on the importance of the structure, the desired analysis type shall be chosen. A simplified pseudo-static analysis is sufficient for less important structure or for the initial design. For a more important structure, a simplified dynamic analysis or a pseudo-dynamic analysis is performed. Further, for most important structures in earthquake prone region thorough dynamic analysis should be performed. 3.4 Input for Design Any analysis can be accurate only when the input parameters are precise. A sophisticated analysis with very poor input data can only lead to inaccurate prediction. In geotechnical engineering, the biggest challenge is to obtain the most appropriate input data. This depends on accuracy of sampling, sophistication of instrumentation, experience of professionals among many. Besides, the uncertainties in earthquake motion can lead to additional problems. Hence, a thorough knowledge of geotechnical conditions at the site of interest is essential. Depending on whether the proposed design is for new or existing structure, appropriate decision on estimating the input parameters is important. 3.5 Output from Design The output from the analysis shall include residual maximum displacement or peak Strain. Depending on the magnitude of this value the design can be revised. For this purpose, the knowledge about the state of structural system at different strain levels is important.

Table 1. Requirements of permanent deformations for different geotechnical structures Type of geotechnical

structure Permissible linear movement Permissible angular

movement Foundation settlement Tilt Retaining wall Linear movement towards fill or

away from fill Tilt

Quay wall Linear movement towards sea Tilt Embankment / earth dam Subsidence & Lateral spread Loss of alignment &

distortion

Slope Horizontal or vertical movement distortion


4. PIANC SEISMIC DESIGN GUIDELINES FOR PORT STRUCTURES Table 2 presents the acceptable level of damage according to PIANC (2001) seismic design guidelines for port structures. The degree of damage is classified in to four categories, namely Degree 1 to 4. The criteria for deciding a particular degree is based on structural and functional aspects. For a smaller degree of damage such as degree 1, both structural and functional aspects should lead to minor or no damage or loss.

Table 2. Acceptable Level of Damage in Performance Based Design Damage Degree Structural aspect Functional aspect

Degree 1 Serviceable Minor or No damage Little / No Loss

Degree 2 Repairable Controlled damage Short term loss of

serviceability Degree 3 Near Collapse Extensive Damage Long term loss of

serviceability Degree 4 Complete Destruction Total collapse Complete loss of

serviceability

Table 2 presents the different performance grades for two different levels of earthquake, namely Level 1 and Level 2. Level 1 earthquake represents the design basis earthquake which has a return period of 250 years and 10 % probability of occurrence in 50 years. That is the most likely magnitude of shaking that a structure may be subjected to in its life span. Level 2 earthquake represents the maximum credible earthquake which has a return period of 500 years and 2 % probability of occurrence in 50 years. A structure might not experience this level of shaking even once. However, the design should be ductile enough to cause destruction with enough warning such that the casualties are minimum. The performance grades are classified in to Grade S, A, B and C. Grade S for superior design and Grade C for low level of importance. Hence, Grade S requires the damage degree to be 1 (serviceable) for both Level 1 and Level 2 earthquake motions.

Table 3. Performance Grades

Performance Grade

Design Earthquake Description

Level I Level II

Grade S Degree 1 Degree 1 Very Important structures

Grade A Degree 1 Degree 2 Primary structures that are difficult to restore if damaged

Grade B Degree 1 Degree 3 Ordinary Structures

Grade C Degree 2 Degree 4 Temporary and easily restorable structures

For the port structures such as quay walls, Japanese guidelines suggest different levels of residual sliding and residual tilting for varying degrees of damage as described in Table 3. Fig 5 explains performance based design of structures schematically with the details of grade type and degree category in accordance with Japanese norms. The vertical axis represents the degree of damage ranging from serviceable to total collapse. The importance of structure increases from left top corner to right bottom corner. The two levels of earthquake motion Level I and Level II are specified in the horizontal scale. Depending on the desired grade of structure, the type of analysis varies from rigorous dynamic analysis for Grade S to simplified pseudo static analysis for Grade C.

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Figure 5. Schematic diagram of Performance grades & Type of analysis (Modified from Iai & Ichii, 1999)

Table 4 Limit for displacement for Port Structures from Japanese Codes

Degree 1 Degree 2 Degree 3 Degree 4

Residual Sliding < 1.5% 1.5 – 5 % 5- 10 % > 10 % Residual Tilting < 3

o 3 – 5

o 5 – 8

o > 8

o

5. DETERMINATION OF PERMANENT COUPLED SLIDING & TILTING

DISPLACEMENTS OF QUAY WALLS The Fig. 6 and Fig. 7 show the various forces acting on the quay wall along with their locations with respect to center of rotation O for un-liquefied and liquefied backfill conditions respectively. The forces and their locations are computed as per the requirements.

Figure 6. Forces acting on quay wall with un-liquefied backfill

θ

O

YFW

FED FBW

FBW

FFW

FFWS

H

Water

α W

ac(x)

ac(y)

h

YBW

YBW

YE

YFW

B

h

Ground Motion

Grade A Simplified Dynamic Analysis

Grade B Response Spectrum

l i Grade C Pseduo Static Analysis

Grade S Rigorous Dynamic

IV–Collapse III–Near Collapse II–Repairable I–Serviceable

L1

Deg

ree

of D

amag

e

Importance of Structure


Figure 7. Forces acting on quay wall with liquefied backfill

With coupled sliding and tilting, the equations of motion include three components, viz.

(1)

(2)

(3)

Solving the above mentioned equations, permanent sliding and tilting displacements are obtained. A useful procedure is adopted to couple the equations for obtaining combined sliding and tilting displacements (Nanjundaswamy, 2009). The newly developed procedure was verified for its accuracy, consistency and applicability to simulate the actual behaviour for which it is designed. For this purpose, the developed model was compared with one model behaviour on shaking table and another field damage (RC 5 quay wall at Rokko Island, Japan during Hyogo-Ken-Nambu earthquake of 1995) where most of the data desired in the analytical work are available. A series of model tests using 1 – g shaking table at S. J. College of Engineering, Mysore on quay wall system. A typical test GH-10 consisted of dense foundation soil, loose backfill and moderately heavy model quay wall subjected to a sinusoidal shaking of around 2Hz at a magnitude of about 4 m/s2 for a period of 12 s. The model quay wall experienced predominantly sliding motion of magnitude 120 mm. Table 6 presents the properties of the model test used as input for the analytical procedure. Fig. 8 presents the pictures of quay wall system after the shaking was completely stopped. It can be observed that the wall suffered a horizontal slide of about 120 mm. The foundation soil, being dense did not experience appreciable damage. However, the backfill soil exerted active earth pressure on the wall which is evident from the extension of backfill soil element by the colored sand grids. Some settlement of the backfill soil and bulging on the front side can also be seen. The results of analysis are presented in Table 7.

YFW

FBLD FBLS

FFW

FFWS

H

Water

α W

ac(x)

ac(y)

h

YBLS YBL

YFW

B

Ground Motion

h

O

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Table 5. Details of input parameters for validation with model test results

Input Parameter Value Wall properties Height (m) 0.3 Width (m) 0.2 Wall splay angle 0 Wall friction angle (degree) 15 Density of wall (kN/m3) 15.2 Height of water table from wall base (m) 0.3 Backfill soil properties Relative density (%) 30 Angle of internal Friction (degree) 34 Backfill slope (degree) 0 Density (kN/m3) 11.3 Foundation soil properties Interface friction angle b/w bottom of wall & foundation (degree) 36 Cohesive stress b/w bottom of wall & foundation (kN/m2) 0

Figure 8: Sliding mode failure of model quay wall after 12 s of shaking

Kobe Port is located in an area 6 km long and 12 km wide and there are two man-made islands, Port Island and Rokko Island. The soils used for landfill were excavated from the Rokko Mountains to the north west of Kobe city. This soil is called locally PI Masado. During the strong shaking of 1995 Kobe (Hyogo-Ken-Nambu) earthquake the waterfront structures at the ports of Kobe suffered substantial damage. Kobe Port had significant ground subsidence as a result of liquefaction during the earthquake. The extent of liquefaction was intense on Port Island and over 250 caissons type quay walls were damaged with a repair cost exceeding US $11 billion (Inagaki et al., 1996). The most severe damage occurred in those quay walls of Rokko Island and Port Island. These walls had been placed on top of gravelly fill consisting of decomposed granite (called locally Masado), which had completely replaced the soft clay layer beneath the wall for improving the bearing capacity and reducing settlements. During the earthquake, the wall top displaced approximately 5 m maximum and approximately 3m average. The walls also settled approximately 1–2m and tilted approximately 4-5 degrees. Caisson type quay walls in Kobe Port including Port Island and Rokko Island were designed using the pseudo-static method, with limit equilibrium mechanics based on the Mononobe–Okabe method using horizontal seismic coefficients ranging from 0.1 to 0.15. The cross-section of a typical quay wall (RC-5, Rokko Island) with its deformation recorded after the earthquake is reproduced in Fig. 6.7 (Inagakii et al., 1996). Table 6.3 presents summary of measured displacements of PC berths at Port Island.

0.120 m


Figure 9. Cross section of quay wall RC-5 in Rokko Island after the Hyogo-Ken-Nambu earthquake of 1995, Japan (Inagaki et al., 1996)

Table 6. Details of input parameters for validation with field damage to quay wall

Input Parameter Value Wall properties Height (m) 16 Width (m) 10 Wall splay angle 0 Wall friction angle (degree) 15 Density of wall (kN/m3) 21 Height of water table from base of wall (m) 14.5 Backfill soil properties Relative density (%) 41 Angle of internal Friction (degree) 36 Backfill slope (degree) 0 Density (kN/m3) 18 Foundation soil properties Interface friction angle b/w bottom of wall & foundation (degree) 40 Cohesive stress b/w bottom of wall & foundation (kN/m2) 0

Figure 10 presents the results from the analytical model for RC 5 quay wall at Rokko Island including the N-S component of recorded ground motion at Port Island Seismograph array. It can be observed that both the components of sliding and tilting are experienced by the wall. The first figure represents the time history of input acceleration (N-S component of Hyogo-Ken-Nambu earthquake of 1995 data) along with the yield acceleration for sliding. The degradation in yield acceleration in about 5 s can be observed. It is interesting to note that the residual portion of the yield acceleration is negligibly small which is also depicted in the permanent displacement which continues till nearly the 20th s. The scenario in case of tilting is different. Though the instant at which the yield acceleration reaches the minimum is nearly the same, the residual portion of yield acceleration for tilting is fairly high. Hence, tilting takes place effectively for less than 5 s beyond which the available strength is good enough to oppose tilting of wall. The final sliding (3.37 m) and tilt (5.96 deg) are comparable with the actual damage to the wall. The final results of validation are presented in Table 7 from both model test and field. It should be noted that the magnitude of sliding from analytical model was comparable with that in the field, but was much higher than that seen from model study. The effect of boundary on the model container could be the cause for such a difference. However, it is interesting that the tilt angles in both the cases almost matched the actual.

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Sliding displacement Tilting displacement Figure 10: Time histories of input acceleration, yield acceleration for for RC 5 quay wall

analysis

Table 7. Comparison of permanent wall displacements from analytical model with those from actual

Sl no Description

Analytical Model Actual

Remarks Sliding (m)

Tilting angle (deg)

Sliding (m)

Tilting angle (deg)

1 Model test 0.251 0 0.120 0 SlidingAnalytical >slidingModel

2 RC 5 at Rokko Island 3.37 5.96

4 m and in some places 5 m

4 – 5 SlidingAnalytical ≈ slidingField TiltingAnalytical ≈ TiltingField

6. CONCLUDING REMARKS This paper provides the methodology for performance based design of geotechnical structures. The design of any geotechnical structure such as foundation, embankment, retaining wall, slope etc. under earthquake loading has sufficient scope to improvise. The present day design guidelines emphasize on force based design which includes evaluating the factors of safety against sliding, overturning and bearing pressure. However, displacement or performance based analysis is being considered to be superior specially for geotechnical structures. Pushover analysis applied to structural frames can provide a basis for assessing the capacity of geotechnical structures under a design earthquake demand. Considering this aspect, it is important to note that the evaluation of precise permanent displacement of geotechnical structure is extremely essential. In this regard, the methodology presented in the paper to estimate the coupled permanent sliding and tilting displacement is useful. It is even more important to validate and show that analytical models are comparable to reality. With such papers, it is hoped that performance based design in geotechnical earthquake engineering takes a big leap in immediate future.

0 40 Time (s)


REFERENCES ATC-40. “Seismic evaluation and retrofit of concrete buildings.” Volume 1 and 2. Applied

Technology Council, California, 1996. De Alba P, Chan C K and Seed H B (1975) “Determination of soil liquefaction

characteristics by large-scale laboratory tests, UCB/EERC-75/14, Berkeley, Earthquake Engineering Research Center, University of California.

FEMA-273. “NEHRP guidelines for the seismic rehabilitation of buildings.” Federal

Emergency Management Agency, Washington DC, 1997. FEMA-356. “Prestandard and commentary for the seismic rehabilitation of buildings.”

Federal Emergency Management Agency, Washington DC, 2000. Hardin B O and Drnevich V P (1972) “Shear modulus and damping in soils – measurements

and parameter effects”, SM&FD, ASCE, V98, SM6, 603 – 624. Iai S (2005) “International standard (ISO) on seismic actions for designing geotechnical

works - An overview”, Soil Dynamics and Earthquake Engineering, Vol. 25:605-615. Iai S and Ichii K (1999) “Performance Based Design for Port Structures”, Proceedings of the

30th Joint Meeting of the U.S.-Japan Cooperative Program in Natural Resources Panel on Wind and Seismic Effects, NIST SP 931, U.S. Department of Commerce.

Inagaki H, Iai S, Sugano T, Yamazaki H and Inatomi T, (1996) Geotechnical aspects of the

January 17, 1995 Hyogoken-Nambu earthquake: Performance of Caisson type Quay Walls at Kobe port, Soils and Foundations (Special Issue):119-36

Nanjundaswamy, P (2009) “A study on Seismic Response of Quay Walls”, Doctoral thesis

submitted to Kuvempu University, India. PIANC (2001) “Seismic Design Guidelines for Port Structures”, International Navigation

Association, A.A. Balkema Publishers, Tokyo. SEAOC (1995). “Vision 2000: Performance-Based Seismic Engineering of Buildings,”

Structural Engineers Association of California, Sacramento, California. Seed H B and Idriss I M (1971) “Simplified procedure for evaluating soil liquefaction

potential”, Journal of Soil Mechanics and Foundation Division, ASCE, 97(9):1249–73.

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75

A Decision Support System for Risk Assessment and Remediation Option

Selection for Contaminated Soils and Groundwater

R. K. Srivastava Department of Civil Engineering, Motilal Nehru National Institute of Technology Allahabad,

email: [email protected] ABSTRACT: Contaminated soils and groundwater are receiving increasing attention due to the greater understanding of their toxicological importance in ecosystem and for human health. Risk assessment based technologies are proving to be useful for selection of remediation technologies for contaminated sites. With a risk based cleanup regimen, sites are remediated to the extent that will render them safe for future land use. With large number of contaminated sites in India, a decision support system would be very helpful in quick, feasible and cost effective remediation process selection. The present study is an endeavour in this direction. The developed DSS has been applied for the contaminated site of UCIL Bhopal and results are reported. Keywords: Decision Support System, Soil/Land Contamination, Soil Screening level, Risk Assessment, Soil/ water remediation technologies

1. INTRODUCTION Soil is one of the most important natural resources that exist on the earth. Soil is an essential part of hydrologic cycle. It contributes towards distribution of surface water, storage of ground water and its recharge. Soil provides sustenance to life nutrients, water, environment, biomass, vegetation, food source and natural habitat of a variety of organisms. In fact the nature of soil plays a decisive role in all aspects of human life. The formation and regeneration of soil is a very slow process. As such, one can consider soil as a non-renewable natural resource. The use of soil for variety of purposes by mankind leads to its degradation. Even though soil has a natural buffering capacity, it can deactivate, filter, store, neutralize and immobilize or act as barrier for a host of contaminants (inorganic, organic, toxic and radioactive) from becoming part of food chain but there is a limit to it. All this is possible and is controlled by various soil properties e.g structure, texture, porosity, cation exchange capacity, pH, Eh and microbial capacity. These properties of soil vary from soil to soil and their response to different contaminants is different. Overall, there is a limit to the capacity of soil to provide protection against contaminants. Therefore, the level of concentration of contaminants with respect to the type of soil becomes most important deciding factor for vulnerability of contaminated soil becoming a waste material and depletion of a precious natural non-renewable resource material.

2. SOIL/ LAND CONTAMINATION Contaminated soil/ land is a term used to describe an area or site which has a high concentration of contaminants that lead to partial or total loss of soil to a level which exceeds the buffering capacity of soils and modifies the properties of soil in a negative way rendering

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its partial or total loss of productivity and other usage. The accumulation of contaminants in soils is a result of usually insensitive human activities. Intensive industrialization generates hazardous wastes comprising of organics, inorganics, heavy metals and munitions. These toxic contaminants, primarily of anthropogenic origin, are broadly classified as metal, non-metal, metalloid, inorganic and organic compounds (Cluis, 2004). Inorganic pollutants may be metals such Ag, Al, As, Be, Cd, Cr, Cu, Hg, Fe, Pb, Sb, Se, Zn and radioactive elements and their derivatives (Meaghar, 2000; Allen, 2002). The principal concern associated with contaminants is toxicity and health risk to humans. The environment contaminated with toxic chemicals, and the by-products are posing a huge task of treatment and/ or management, in a safe manner. It is therefore essential to contain or mitigate these organic and inorganic contaminants so as to prevent them from contaminating surface and ground water and/ or becoming part of food chain. Heavy metals in soils are receiving increased attention due to the greater understanding of their toxicological importance in ecosystems and for human health. Heavy metal contaminated soils can be a long-term environmental concern and a potential financial liability to landowners. Hence, the assessment of heavy metal contaminated soil has received much attention in the last few decades. Land contamination has become a major environmental issue following the rapid industrial development that has taken place in many parts of the world in recent years. Elevated concentrations of heavy metals in soils are potential long-term environmental and health concerns because of their persistence and cumulative tendency in the environment, and their associated toxicity to biological organisms (Nriagu and Pacyna 1998). Furthermore, restricted use of contaminated lands and the costs of soil remediation also pose liabilities and financial burdens on landowners and other stakeholders. As a consequence, environmental assessment of lands with respect to heavy metal contamination and identification of its environmental and health implications have become increasingly important in geoenvironmental research (Balaramudu 2007). 2.1 Problems and Importance: Contaminated Sites

• The pollution is not visible: it is located in and moves through the soil. • Its development and movement can be slow: several years or even decades. • The pollution is only discovered when it reaches a ‘target’, e.g. drinking water supply,

fruit or vegetable (and even here it is not noticed immediately). • The consequences can be serious: accidents, illness, degradation or even destruction of

water resources or flora and fauna. • The high social cost of damage to human health and loss of land. • Reduced land values. • Loss of land for agriculture and urbanization. • The serious questions of political accountability is raised when they are discovered. • The important financial resources are monopolized for treatment, to the detriment of

development. • The high cost of their treatment to industry.

2.2 Soil Contamination- Indian Scenario (Balaramudu 2007) Cadmium concentration in some hair samples of people living near dumping site in India exceeded the level associated with learning disorder in children. Levels of most of the trace elements in hair were significantly higher in dumping site than those in control site in India,

Risk Assessment and Remediation Option Selection for Contaminated Soils and Groundwater 77

suggesting direct or indirect exposure to those elements from dumping wastes (Agusa et al., 2003). Ackland (2006) reported that approximately 30,000 hectares of India land soils are contaminated by toxic heavy metals lead, arsenic, mercury and cadmium produce from thermal power plants that produce fly ash.

Patel et al. (2005) reported the arsenic contamination in water, soil, sediment and rice samples of central India (Ambagarh Chauki, Chhattisgarh). The soils have relatively higher contents of arsenic and other elements. The mean arsenic contents in soil of this region are much higher than in arsenic soil of West Bengal and Bangladesh. The lowest level of arsenic in the soil of this region is 3.7 mg/kg with median value of 9.5 mg/kg. The arsenic contents in the sediments are at least 2-folds higher than in the soil.

An environmental geochemical investigation was carried out by Krishna and Govil (2004) on soils in and around the Pali industrial development area of Rajasthan to determine the Pb, Cr, Cu, Zn, Sr and V contents. Levels of the these metals in soils around the industrial area were found to be significantly higher than their normal distribution in soil such as Pb-293 mg/kg, Cr-240 mg/kg, Cu-298 mg/kg, Zn-1,364 mg/kg, Sr-2,694 mg/kg and V-377 mg/kg. The soil in the industrial area of Hyderabad, India has been highly polluted with heavy metals like chromium, vanadium, cadmium, and barium. Studies carried out by the National Geophysical Research Institute (NGRI) have revealed that levels of the these metals in soils were found almost three times the normal level such as barium of 1,300 mg/kg, cadmium of 100 to 300 mg/kg, chromium were between 50 to 150 mg/kg and 1400 mg/kg of strontium were found. The phosphorite rocks of the Roorkee area of India contain cadmium. The area under study has soil rich in cadmium- containing minerals, as a result of which water is likely to leach out mobile cadmium and the ground water and crops could become contaminated. This exposure to cadmium is probably causing ailments so far unreported (Khwaja et al. 1997). In alkaline soils of the Punjab, cadmium was retained by adsorption on the mineral interface and by interaction with organic matter and calcium carbonate. 3. SOIL REMEDIATION TECHNOLOGIES

Table 1 summarizes various remediation technologies along with their advantages, disadvantages and applicability.

Table 1. Remediation Technologies METHOD DESCRIPTION ADVANTAGE DISADVANTAGE

Mechanical separation

Gravitational separation Sieve Analysis Magnetic separation

Significant volume reduction of contaminated soil.

not satisfactory decrease in volume in case of homogenous distribution of Pollutants. The separated part of the contaminated soil must be cleaned up using another method.

Contd……..

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Electro- remediation

Based on pollutant migration in an electric field. Migrating particles should have permanent electric charge or have to be polarized, so the technique is used to remove heavy metals or polar compounds.

Only method for in-situ removal of heavy metals from contaminated soil. Applicable to all metals.

Any heterogeneity of the soil body decreases the effectiveness of the method. Considerable acidification of the remediated soil is side effect of this method.

Cofferdam system

A system of barriers is made from different substances placed under the soil’s surface. Different chemical substances present in the barriers, convert pollutants into environmentally friendly forms.

Applicable to both organic, inorganic contaminants.

Any damage to the barrier decreases its efficiency The underground water flow direction must be well known.

Pump and Treat Method

Pump and treat involves pumping out contaminated groundwater with the use of a submersible or vacuum pump, and allowing the extracted groundwater to be purified by slowly proceeding through a series of vessels

Useful for removing contaminants when a number of bore holes are to be made on the contaminated site. Removes groundwater contamination with a variety of dissolved materials. Cost effective

Difficult to reach sufficiently low concentrations. Not applicable for soil media.

Soil washing

Used for removing inorganic contamination, such as heavy metals, radio-nuclides, toxic anions and others.

Highly effective method for cleaning up strongly contaminated soils.

High costs of construction of the cleaning installation and utilization. Generation of large amount of solid and liquid wastes that need later management.

Ex-situ immobilization of contaminants

Used for the neutralization of organic and inorganic compounds in soil. For neutralization purposes, the contaminants’ bonded substances are added (e.g. cement), which completely blocks the pollutants in the soil.

Fast and easily applicable method Relatively low costs of investment and operation

Generation of significant amount of solid wastes High invasivity to the environment

Chemical and photochemical reduction

Allows the total mineralization (by chemical reactions) of organic contaminants or the effective transformation of organic and inorganic contaminants into non-toxic, less toxic or chemically inert forms.

Wide range of applications & Low costs of operation

High invasivity to the soil and the environment.

Soil flushing

This method is similar to soil washing and is used for the same group of pollutants (heavy metals), but is applied in situ.

Lack of solid wastes. Applied in situ, without soil removal.

A large amount of liquid and semi-liquid wastes are generated. Incomplete removal of contaminants (strongly bounded heavy metals remain in the soil)

Contd……..


Composting

Contaminated soil is explored and stored as pile or thin layer distributed over larger area, for the degradation of contaminants. Some organic contaminants (oil origin compounds, non-halogen compounds and some halogen compounds and pesticides) removed from the soil, by way of biological degradation.

Specialist equipment, process controlling and trained staff not required.

Not effective for strongly contaminated soils. A considerably large area for storage is required. Time taking process

Biological filters and bioreactors

Based on the biological activity of microorganisms. The contaminated soil is mixed with water and as a suspension is moved into reaction chamber microorganisms removes the contaminants as result of sorption and/or transformation.

Effective and relatively fast remediation technique (the fastest biological technique). Soil retains its properties and could be replaced on the reclaimed site.

Construction of special installation is required. Large amounts of wastes (solid, liquid) are generated.

Bioremediation

Based on the physiological activity of microorganisms. Organic compounds are used by microorganisms as substrates in energetic processes. This method is used to clean up soils contaminated by organic compounds and heavy metals. (e.g.Bacterium Deinococcus radiodurans can be used)

No special installation or trained staff required. Almost non-invasive to the environment.

Not applicable for highly contaminated soil. Method efficiency is highly dependent on weather and climatic conditions (low temperatures and humidity decreases the efficiency of the method).

Phytostabilization

This process is based on the ability of roots to immobilize pollutants. The process takes place on the surface of roots as an adsorption effect. Contaminants are absorbed into roots and precipitated in the roots’ area.

Low-cost method. Positive influence on the environment, as an effect of the reconstruction of plant cover on the soil surface.

Contaminants are not removed from the soil but only immobilized. The applied plants usually require intensive fertilization

Phytoextraction

In this method, contaminants are picked up by the roots of plants and transported to their over ground parts, and then removed together with the crops.

Low technical equipment requirements.

In case of soil heavily contaminated by heavy metals, a phytotoxic effect can occur and make the process difficult to conduct. There is a risk of animals bringing contaminants into the food chain.

Plant cover

Long-term, self-standing systems of cultivated plants that are introduced to the danger surface area. Plant cover can reduce the danger to an acceptable level and requires minimum conservation.

Prevention and minimization of surface erosion by forming a self-sufficient ecosystem. Supports the biodegradation of contaminants in soil.

Necessity for long-term monitoring and maintenance of plant cover. Contaminants can be incorporated into the food chain through plants.

http://en.wikipedia.org/wiki/Deinococcus_radiodurans





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4. RISK ASSESSMENT The objectives of Risk Assessment are:

• To establish baseline risks and whether site remediation or other action is necessary • To determine a tolerable level of contaminants that can remain in place with adequate

protection of public health • To enable comparison of potential health impacts of various remediation techniques

It is a technique by which the actual or potential adverse effects of contaminants on plants, animals, or ecosystem can be assessed in a systematic manner. It involves: The assessment of the potential for exposure to contamination; & The severity of the effect of such exposure It identifies: The potential for exposure to exceed the determined maximum allowable dose or allowable exposure. Its purpose: To protect human health & the environment from current & potential threats posed by uncontrolled hazardous substances releases. It is the process of evaluating alternative actions, selecting options and implementing them in response to risk assessment. The objective of risk management is to ensure that the risks associated with a contaminated site are properly managed. In present context basically risk describes the combination of the probability and the effects of contamination, like adverse environmental effects on human health, on ecosystems, or on water resources. To assess risks at a site, the following information is required:

• Contaminants and their magnitudes • Extent of contamination • Toxicity of the contaminants (carcinogenic or non-carcinogenic) • Possible ways in which exposure may occur.

4.1 Quantification of Risk Screening Levels (DOE 2003) It is the process of identifying and defining areas, contaminants, and conditions, at a particular site that do not require further attention for remediation. Basis for determination of Screening Levels

• Risk : It can be categorized as carcinogenic and non-carcinogenic risks • For carcinogenic contaminants, Risk = Chronic Daily Intake * Slope Factor, where

Slope factor is the maximum daily intake for a particular contaminant that is likely to be without an appreciable risk of developing cancerous cells within human body

• For non-carcinogenic contaminants, risks are defined as Hazard Index (to differentiate from carcinogenic risks) Hazard Index = Daily Intake / Oral Reference Dose(RfDo), where RfDo is the max. daily intake without appreciable risk of deleterious health effects


• Both Carcinogenic Risk and Hazard Index are unitless probabilities of an individual either developing cancer or a deleterious health effect as a result of exposure to contaminants which have been limited to 1*10-6 and 1 respectively by the World Health Organisation.

Equations for Ingestion of Non-carcinogenic Contaminants in Soil THQ (target hazard quotient) unitless BW (body weight) kg AT (averaging time) yr EF (exposure frequency) d/yr ED (exposure duration) yr IR (ingestion rate) mg/d RfDo=oral reference dose (mg/kg-d). chemical-specific Averaging Time (AT) = Exposure Duration (ED) for non-carcinogenics Ingestion of Carcinogenic Contaminants in Soil - Age Adjusted

Ingestion of Carcinogenic Contaminants in Soil – Non Adjusted TR (target risk) unitless BW (body weight) kg AT (averaging time) yr EF (exposure frequency) d/yr ED (exposure duration) yr IR (ingestion rate) mg/d Screening Level Equation for Migration to Ground Water

IREDEFmgkgRfDo

yrdayATBWTHQkgmgSL***)/(610*/1

)/(365***)/(−

=

adjustedageo IFEFmgkgSFyrdayATTRkgmgSL

−−=

**)/(10*)/(365**)/( 6

)317(

)317()317(

)61(

)61()61( **yrsage

yrsageyrsadjage

yrsage

yrsageyrsadjageadjsoil

BWEDIR

BWEDIRIF

−

−−−

−

−−−− +=

IREDEFmgkgSFyrdayATBWTRkgmgSL

o ***)/(10*)/(365***)/( 6−=

5.1)]'3.0([ HKCSL dw ++

=

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Cw (target contaminant concentration in soil)mg/L = Health based limits x dilution factor (may be calculated or set to site-specific default ) : default value 20 H' (dimensionless Henry's law constant) - chemical specific Kd (soil-water partition coefficient) L/kg = Koc x 0.002 (organics) -chemical specific Koc (soil organic carbon/water partition coefficient) L/kg - chemical specific Table 2 Provides Equation values for calculation of Screening Levels

Table 2. Equation Values

Non-Carcinogenic Parameter

Value Carcinogenic Age-adjusted Parameter

Value Carcinogenic Non Age-adjusted Parameter

Value

Target Hazard Quotient (unitless)

1 Target Risk (unitless) 1.0E-6 Target Risk (unitless) 1.0E-6

Body Weight (kg) 15 Adult Body Weight (kg) 70 Body Weight (kg) 70

Child Body Weight (kg) 15

Exposure Duration (yr)

6 Adult Exposure Duration (yr)

24 Exposure Duration (yr) 25

Child Exposure Duration (yr)

6

Exposure Frequency (day/yr)

350 Exposure Frequency (day/yr)

350 Exposure Frequency (day/yr)

250

Ingestion rate (mg/day)

200 Adult Intake Rate (mg/day)

100 Ingestion rate (mg/day) 50

Child Ingestion Rate (mg/day)

200

Average Lifetime (yr) 70 Average Lifetime (yr) 70

Age-adjusted Ingestion Factor (mg-yr/kg-day)

114.29 Dilution Factor 20

5. DEVELOPMENT OF DECISION SUPPORT SYSTEM The Decision Support System (DSS) is developed based on the following four criteria:

• Type of media: Whether contamination is present in soil or ground water or both. Methods like electro-remediation and mechanical separation are suitable for soil only while pump and treat method and impermeable barriers are suitable for water only.

• Nature of contaminant: Whether the contaminants belong to heavy metals, organics & pesticides or other inorganics. Methods like soil flushing and chemical precipitation are more suitable for organics and pesticides while ex-situ immobilization of contaminants and chemical oxidation are suitable for inorganics only.

• Health risk: The health risk due to various contaminants have been calculated using world health organization (WHO) standards for severity of different chemicals. WHO has classified all the chemicals on the basis of risk to health in three categories viz. Terminal and incurable diseases (leading to death of human being), curable diseases not involving risk to life of people, and concentrations upto which minor or no health risk is associated. Based on these three classes, three categories have been defined in the decision support system i.e. severe health risk involving risk of death, moderate


health risk and on health risk. • Concentration of Contaminants: Some methods are suitable only for bringing down

the concentration of less contaminated soil/ groundwater and vice-versa.

5.1 Decision Support System A java based decision support system has been developed which provides suggestions for remediation option for the contaminated site. The application design is given in Table 3.

Table 3. Application Design INPUT PROCESSING OUTPUT

pH of the soil and groundwater

Contaminated media Type of land use Prominent

contaminants at the site

Contaminants and their concentration

Area of the contaminated site

Comparing the concentration of each contaminant with the corresponding Screening Levels(SSL) for the given land use type & accordingly deciding the necessity of remediation Factors governing application of a methodology:

1. Type of contaminants at the site: Heavy metals, organics, other inorganics, etc.

2. Effectiveness of the method 3. Methodological Constraints 4. Cost incurred in application

Most feasible Methodology (s) along with their advantages, disadvantages and relative cost

5.2 Framework of Decision Support System The inputs: area, land use, contaminants and pH of soil are obtained from the user. Next step is to get the concentrations of these contaminants. Now these concentrations are compared with their respective soil screening levels. If the concentration of contaminant is less than its soil screening level then there is no requirement of any remedial measures and the site is safe for this contaminant. But if the concentration of contaminant is greater than its soil screening level remediation will be required. It will be required to reduce the contaminant concentration upto a desired low level based on the soil condition and its future uses. The remedial technique is then decided based on various factors viz. type of contaminants, effectiveness of remedial methods, time taken by the method to reduce the concentration of contaminant to required level and cost incurred in applying the method. The most appropriate methods are given as the output of decision support system. 5.3 Elements of the Database The database development consists of four parts:

• Data punching of screening levels: Around 750 chemicals have been considered in this decision support system whose raised concentration leads to contamination of soils or ground water.

• Feeding the methodologies for application to contaminated media: Different methodologies that are worldwide used for remediation of contaminated soil and ground water have been punched in this database. The techniques have been punched based on their effectiveness, applicability (applicable to heavy metals, inorganics, biological organics etc), pH range in which they are effective etc.

• Feeding health risk based concentrations for the contaminants: These are the desired

84 ETGE 2012

level upto which the concentration of contaminants has to be reduced for no health risk.

• Applicability conditions and constraints for different methodologies: These are the elimination criteria for remedial techniques based on pH, area of site, effectiveness of method being applied, time taken by the given method to reduce the contaminant level to desired concentrations.

5.4 Selection Criteria of Remediation Technology The criterion for selection of a remediation technology is based on various factors including:

• Effectiveness of the remediation technique • Applicability of methodology • Availability of a given remedial method on a given site • Operational requirements of various remedial measures • Limitations of remedial techniques • Cost of applying a given methodology • Time taken by a given technique to reduce the contaminant level to desired level

6. CASE STUDY The developed decision support system has been used to obtain remediation plan for the abandoned UCIL premises at Bhopal, Madhya Pradesh. Site Overview

• M/s. Union Carbide India Ltd. (UCIL), manufactured pesticides and the associated intermediate chemicals at their Bhopal unit from 1969-1984. The unit was closed down in December 1984 as a result of the infamous accident of leakage of methyl iso-cyanate gas.

• The solid, semi-solid and liquid wastes generated during the manufacture of pesticides and associated chemicals were dumped by UCIL within their premises from 1969 to 1984.

• The unscientific disposal of these wastes could have resulted in contamination of land and water environment in and around the UCIL premises and require remediation.

• Monitoring of groundwater collected around UCIL premises indicates isolated contamination in the UCIL premises.

It is estimated that the amount of contaminated soils is approximately 6,50,000 m3 and major contaminants are heavy metal and pesticides. Table 4 provides list of selected contaminants and their screaming levels. Since clean-up standards for hazardous waste contaminated sites are yet to be developed and notified by the regulatory agencies in India, the latest (2009) standards/screening levels published by USEPA , December 2009 have been used presently. - National Implementation Plan, NEERI - 2010 6.1 Selection Criteria for Remediation Methodology Purpose: To identify and then evaluate remedial methods (or combination of methods) which may be suitable for use on UCIL site.


Table 4. Screening levels for the contaminants

S . No. Contaminant

Screening Levels (Concentrations detected)

Industrial Soil (mg/kg)

Ground water protection (mg/L)

1 Mercury 34 (ND) 0.03 (0.61)

2 Aldicarb 620 (940) 15 (ND)

3 Naphthylamine 9.6E-01 (181.859) 3.3E-02 (ND)

4 Zinc 31000 (3.756) 5.0 (6.1345)

5 Cobalt 300 (ND) 4.7 (8.4015)

6 Carbaryl 620 (1100.25) 140 (ND)

Prerequisites of a Remedial Method:

• Applicable to the contaminant and media o (e.g. soil, construction, debris, ground water etc.)

• Effective in achieving specified contamination related objectives • Practically feasible • Cost effective

Feasible: Composting - contaminated soil is stored as pile or thin layer distributed over larger area, for the degradation of contaminants. Construction of physical/permeable barriers: contaminant present from a long time, etc. Table 5 presents facts and inference from site analysis.

Table 5. Site Analysis Fact Inference

Contaminants persisting at the site from a very long period

In-situ technologies (thermal desorption, permeable reactive barriers, bioremediation, etc. ) not feasible. Ex-situ treatment methodologies may be used.

Presence of organic contaminants

Mechanical separation, electro remediation, compositing, etc. can not be used.

High cost Bioreactors, biological filters, soil flushing, cofferdam, etc. cannot be used

Presence of heavy metals Biological methods (Bioremediation, phytostalization, etc.) not much helpful. *Can be used after the treatment for long term improvement of soil quality

Table 6 presents selection of remediation process for the present case of contaminated land and ground water at UCIL, Bhopal, using the developed decision support system.

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Table 6. Selection of Remedial Process Suggested Methods of Remediation

Area Suitable Method Advantages

Soil Contamination

Storage and extraction (landfill)

Short time of excavation of contaminated soil Does not require highly specialist equipment Economical Can be used for both organic and inorganic contaminants

Groundwater Pump and treat system

Useful for removing contaminants when a number of bore holes are to be made on the contaminated site. Removes groundwater contamination with a variety of dissolved materials. Cost effective.

7. CONCLUDING REMARKS

After analysing the contamination at the site, the remediation strategies suggested for application are to have a Pump and Treat system for the remediation of groundwater and to make an onsite landfill for reducing the contaminant concentrations in the soil. Cost estimates for establishment of onsite secured landfill: Based on the data/information available on landfill constructions, the cost of construction of the landfill facility varies from Rs. 600 to 900 per m3. Thus, the cost of construction of secured landfill will be in the range of Rs 78 crore to 117 crore (approx. Rs. 100 crores ). Cost estimates for remediation of contaminated groundwater: The capital cost for such pump and treat unit shall be in the range of Rs. 25 to 30 lakhs. The operating and maintenance cost of such unit is in the rage of Rs. 10 to 15 lakhs per annum including cost of activated carbon and its disposal. REFERENCE Ackland, L. (2006). News Release on Deakin doctors revive sick soils,

www.deaking.edu.au/news/upload. Agusa, T., K, Nakashima, T.E., Minh, T.B., Tanabe, S., Subramanian, A. And Viet, P.H.

(2003). Preliminary studies on trace element contamination in dumping sites of municipal wastes in India and Vietnam. Journal dePhysique IV France, Vol:107, p.21.

Allen, H.E (2002) Bioavailability of metals terrestrial ecosystems; Importance of

partitioning for bioavailability a invertebrate, microbes, and plants. Society of Environmental Toxicology and Chemistry (SETAC), Pensacola, Fla.

Balaramudu, P. (2007). Electrokinetic remediation of contaminated fine grained soil. Ph.D Thesis submitted at MNNIT Allahabad, August, 2007

Cluis, C. (2004). Junk-greedy greens: Phytoremediation as a new option for soil decontamination. BioTeach, Vol.2, pp.61-67

ENVIRONMENT AGENCY (2003). CLR11 Model Procedures for the Management of Contaminated Land.

Khwaja A.R., Singh R., Raju M and Tandon S.N (1997). The geoenvironmental cycle of cadmium: a case study, The Environmentalist, Volume 17, Number 2, 1997, pp 103-

http://www.deaking.edu.au/news/upload


108 (6). Krishan, A. K. and Govil, P.K. (2004). Heavy metal contamination of soil around Pali

Industrial Area, Rajasthan, India. Environmental Geology, Vol. 47, No. 1, pp. 38-44. Meaghar, R.B. (2000). Phytoremediation of toxic elemental and organic pollutants.

Current Opinion in Plant Biology, Vol. 3, pp. 153-162 National Geophysical Research Institute data on contamination present at abandoned UCIL

site at Bhopal, Madhya Pradesh. Nriagu, J.O. and Pachyna, J. M. (1998). Nature (London), Vol. 333: pp. 134-149. Patel KS, Shrivas K, Brandt R, Jakubowski N, Corns W, Hoffmann P. (2005). Arsenic

contamination in water, soil, sediment and rice of central India. Environ Geochem Health. Apr.,27(2), pp. 131-45

USEPA (2009), Screening levels. World Health Organisation guidelines on Chronic Intake levels.

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89

Risk Analysis of Bearing Capacity of Shallow Foundations

S.M. Dasaka

Department of Civil Engineering, IIT Bombay, Mumbai – 400 076, India email: [email protected]

ABSTRACT: Field load tests are regarded as the most suitable technique to obtain load-settlement response of foundations resting on soil/rock, and find its place in several codes of practice worldwide. The load-settlement data obtained from plate load tests is routinely used in the evaluation of several design parameters of soil, such as bearing capacity, settlement of foundations, modulus of subgrade reaction, Young’s modulus, etc. However, these tests often produce unrealistic results, due to various sources of uncertainties in testing procedures, and may create chaos in the design decisions, if not properly accounted for. In this paper, a failed field plate load test is discussed and various uncertainties associated with the field testing are identified. The effects of these uncertainties on the load-settlement behavior are systematically analysed, through a series of laboratory plate load tests, simulating the field conditions. By comparing the results obtained from the laboratory and field load tests, it is observed that the bearing resistance of the soil is highly overestimated due to non-maintained load testing. Based on laboratory load test data, resistance factors are derived for each stage of loading, which when used with non-maintained load test data, can partially compensative the error associated with this uncertainty on the load-settlement response. Keywords: Bearing capacity, codal provisions, Plate load test, maintained load, load-settlement response, uncertainty, resistance factors 1. INTRODUCTION Risk analysis is a design tool to explicitly consider the effect of various sources uncertainties on the performance of a system. Geotechnical engineering is considered as a good example of an engineering discipline where every stage of it invariably encompasses several uncertainties. Uncertainties in geotechnical engineering may arise from using a geological model developed for a site based on information obtained from limited borelogs, use of soil properties measured/interpreted through in-situ and laboratory testing, use of simplified transformation models for bearing capacity and settlement predictions, to name a few. Keeping in mind the scale of these uncertainties effecting the foundation design, it is rightly said that the subject of foundation engineering is an art and science of molding materials we do not fully understand into shapes we can not precisely analyze to resist forces we can not accurately predict, all in such a way that the society at large is given no reason to suspect our ignorance (Coduto, 1994). Field plate load test is considered as one of the most suitable techniques to obtain load-settlement response of foundations resting on soil/rock, due to high degree of uncertainties associated with other in-situ and laboratory testing procedures, and errors due to simplified transformation models developed for estimation of bearing capacity and settlement of shallow foundations. Table 1 summarizes the results from a prediction symposium on behavior of shallow foundations held at Texas A & M University, USA. The predicted bearing capacities of footings resting on a medium dense, fairly uniform, silica sand deposit, are presented for 5 different footings, and compared with the load test results. The wide scatter of predicted bearing capacities may partly be attributed to the simplified models

90 ETGE 2012

adopted (Briaud and Gibbens, 1999). Similarly, errors in field and laboratory testing, and statistical errors due to limited sampling may also result in greater differences between the measured and predicted performance of foundations. Figure 1 shows the effect of uncertainty on the performance of a shallow foundation. It can be noted from the figure that the footing with higher factor of safety in fact exhibits a higher probability of failure, and a realistic assessment of the safety of the foundation is only possible if the effect of various sources of uncertainty is properly accounted for in the foundation design process. The past three decades of research in the area of geotechnical risk and reliability identified that the conventional factor of safety approach alone can not quantify the risk associated with various foundation design decisions, and strongly emphasized the need for a systematic study of various sources of uncertainty, and quantification of their influence on the probability of unsatisfactory performance of a foundation system. An attempt is made in this paper to demonstrate the level of uncertainty associated with a field plate load test, and its effect on the load-settlement response. The load-settlement data obtained from plate load tests is routinely used in the evaluation of several design parameters of soil, such as bearing capacity, settlement of foundations, modulus of subgrade reaction, Young’s modulus, etc. This information is very much essential and assists the design engineers in the decision process, and to arrive at a suitable foundation for a structure, taking into consideration safety, serviceability and economic aspects. In this paper a major error associated with conducting a field plate load test is described, and its effect on the load-settlement behavior is quantified through simulation of field conditions in the laboratory. It is common practice to conduct field plate load test under reaction loading, as it is a simple loading method compared to other alternatives, such as gravity loading. In this, a series of incremental loads is applied through a hydraulic jack working against a reaction loading in the form of a Kentledge assembly/anchor piles, with a definite time lag between consecutive load increments. These loads on the plate shall be applied in cumulative equal increments of 100 kPa or one-fifth to one tenth of the estimated bearing capacity (IS: 1888-1982, ASTM D1194-94). IS 1888, in the case of clayey soils, suggests that the load should be increased to the next stage either when the curve indicates that the settlement has exceeded 70 to 80 percent of the probable ultimate settlement at that stage or at the end of 24 hours period. For granular soils, each load increment should be kept for not less than one hour or up to a time when the rate of settlement gets appreciably reduced to a value less than 0.02 mm/minute, and there is no mention on maintaining the load constant during loading stages. However, ASTM explicitly specifies that after the application of each load increment, the cumulative load be maintained for a selected time interval of not less than 15 minutes. Such shortcomings in the codal provisions may prove disastrous, if the codal provisions are followed blindly, without insight. Moreover, due to various sources of uncertainties in testing procedures, viz. lack of suitable equipment and experimental know-how, inadequate codal provisions, these tests often produce unrealistic results, and create chaos in the design decisions.

2. MOTIVATION AND AIM For jack-reaction type of loading, with application of load on the plate, the plate penetrates into the soil with a corresponding reduction in the applied pressure. In principle, the load on the plate should be maintained constant during each loading stage.

Risk Analysis of Bearing Capacity of Shallow Foundations 91

Table 1. Measured and predicted bearing capacity of shallow foundations (Briaud and Gibbens, 1999)

Figure 1. Effect of uncertainty on the failure probability (Lacasse, 2001) However, in the routine practice, due to negligence, lack of technical know-how of the crew, and inadequate testing equipment, this condition is seldom satisfied, and goes unnoticed and not reported elsewhere. This research work is emanated from the recent experience of the first author at a power plant site, where few field plate load tests were conducted, for which he played an advisory role. The details of the field plate load test, and observed major deviations from the standard code of practice are delineated in the following sections. Few well controlled laboratory tests are conducted to understand the effect of maintained and non-maintained load tests on the load-settlement response. The effect of the above uncertainty on the load-settlement response is analysed, and a set of resistance factors are derived for each load. These resistance factors, when multiplied with the corresponding observed maximum settlement, can partially compensate the error associated with using the conventional hydraulic pumping units, which cannot hold the applied pressure during the loading stages. 3. CASE STUDY A series of plate load tests were conducted at a Power plant site as part of in-situ studies for selection of a suitable foundation system for supporting important plant structures. The plate load tests were required to be carried out as per the specifications outlined in Indian Standard Code of Practice (IS 1888:1982). The size of the plate selected was 0.6 m x 0.6 m, and the test was conducted at 4 m depth below the natural ground level. No ground water table was observed in the vicinity of the plate, as the test was conducted during a post-monsoon period. The load was applied on the plate through hydraulic jacking against a heavy kentledge of sand bags, in 5 equal load increments of 100 kPa. A pressure gauge is attached to the hosepipe connecting manually operated pumping unit and the hydraulic jack, to control the pressure

92 ETGE 2012

applied to the plate assembly. The pumping unit was operated manually at the start of every load stage, till a corresponding pressure was reached in the pressure gauge, and left unattended till the next loading stage. For the hydraulic jack used in the present study, to apply a pressure of 100 kPa on the plate of 0.6 m x 0.6 m, the pressure of the oil in the hosepipe should correspond to 80 kg/cm2. Settlement of plate was measured using two manually recording dial gauges placed on the plate at diagonally opposite locations. It was noted during every loading stage, that there was a drop in the pressure on the plate, as the plate settles into the soil. The rate of decrease of pressure on the plate may be directly proportional to the rate of settlement of the plate. In order to illustrate the above effect, 4th loading stage was considered. A pressure of 400 kPa was initially applied on the plate, with the corresponding pressure gauge reading of 320 kg/cm2. If the pressure on the plate was maintained constant, the pressure gauge reading (320 kg/cm2) must be constant throughout this loading stage. However, due to the fact that the hydraulic jack was not able to hold the applied pressure constant, a reduced pressure gauge reading of 250 kg/cm2 was noticed (Figure 2), after almost 12 hours from the application of 4th load increment, which would correspond to a pressure of 312.5 kPa on the plate. A similar drop in the pressure gauge reading was noticed with elapsed time at all the loading stages. Adding a simple servo-controller or any other pressure holding device to the pumping unit, could be a possible solution to maintain the load on the plate constant during each loading stage, and obtain reliable load-settlement data.

Figure 2. Observed pressure drop during 4th loading stage

4. OBSERVED UNCERTAINTIES DURING FIELD TESTING The following are some of the major uncertainties noticed during the field plate load test.

1. The pressure on the plate was not maintained constant during loading stages 2. The supports of the datum bar carrying the dial gauges were placed very close the

edges of the plate, and rested in the influence zone. 3. The influence zone surrounding the plate was highly disturbed due to movement of

crew, and placement of testing related accessories. 4. The positioning of the dial gauges were wrong

5. EXPERIMENTAL PROGRAMME To demonstrate the effect of the observed loading mechanism on the load-settlement behaviour, two types of laboratory plate load tests are conducted on remolded marine clay, obtained from a site in the west coast in Mumbai. In the first test, the load on the plate during each loading stage is maintained constant and in the second test the load at each stage is not maintained, simulating the observed loading behavior, obtained by using a conventional


hydraulic manual pumping unit. The comparison between the two test results are reported in terms of time-settlement and load-settlement response, and the details are presented below. 5.1. Test bed preparation The clay bed for all the tests is prepared in a test tank of plan dimensions 46 cm×46 cm and 41 cm in depth. The inner surface of the tank is coated with metal paint to reduce the boundary effects. Apart from this, a fine thin layer of oil is also applied to its internal walls. The moist soil in the tank was compacted in layers and for each layer 9 kg of soil having water content 30% is used. Each layer of soil is then compacted with a special hammer with equal number of blows so as to achieve clay bed with uniform density and consistency. 5.2. Load tests For the load maintained tests, the tank with clay bed is mounted on a reaction frame and a lever arm based incremental loading, which helps to maintain the load during each increment of loading, is used. A plate of size 10 cm×10 cm×0.7 cm is kept at the centre of the tank and then load is applied by keeping appropriate load on the lever arm to get the desired load on the plate. The load was maintained during each loading stage for a duration of 1 hour. Settlements are observed with the help of two diagonally placed LVDTs at 1, 2.25, 4, 6.25, 9, 16, 25, 36, 49, 55 and 60 minutes from the start of each loading stage. A typical view of the test setup and the reaction frame is shown in Fig. 3(a), and Figure 3(b) shows a close view of the arrangement of LVDTs, load cell, and plate arrangement.

(a)

(b)

Figure 3 (a). Laboratory plate load Test setup with lever arm mechanism for maintained loading; (b). Detailed view of the load cell and plate arrangement

To simulate non-maintained load tests, the tank was mounted on a self-supporting reaction frame and load is applied on the plate through a hand operated pumping unit-hydraulic jack assembly. The load during each loading stage is applied by operating the hand held lever of the pumping unit, till the corresponding pressure is reached in the pressure gauge, and unattended until the next load increment. The displacements are continuously recorded with the help of LVDTs. The duration of each loading stage is kept as 1 hour, similar to the case of load maintained tests. Figure 4 shows the loading mechanism, and arrangement of LVDTs, load cell, and plate arrangement.

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(a)

(b)

Figure 4 (a). Laboratory plate load test setup with hydraulic jack-reaction load mechanism for non-maintained loading; (b). Detailed view of the load cell and plate arrangement

5.3. Results and discussion The time vs. load response is plotted for both the cases, described in the above section, and shown in Figure 5. It is obvious that a stepped pattern is observed when the load is maintained constant at each loading stage. For the case of non-maintained load, upon each load increment a gradual and steep reduction in applied load is observed, with an associated settlement of plate.

Figure 5. Variation of pressure on the plate with time Figure 5 shows the time versus settlement profiles for both loading cases. In the case of maintained loading, the settlements increased rapidly upon loading and reach a rate less than 0.02mm/minute, within an hour of loading, for each loading stage. Whereas for non-maintained loading, after a high initial settlement a rebound of plate is observed, which gradually increased with time and finally attains a near constant value. A typical load vs. settlement response is shown in Figure 6 for both loading cases. It is clearly seen from the figure that the load-settlement response for both the cases is quite different. Based on the above observations, it can be noted that the load tests with conventional hydraulic pumping units, which cannot hold the pressure constant during loading stages, lead to flatter load-settlement response. The data obtained from such tests may overestimate the bearing capacity and underestimate the settlements corresponding to any load on the plate.


Figure 5. variations in settlement profiles with time

Figure 6. Observed pressure-settlement response

By visualizing the load-settlement response observed in the laboratory, it is prudent to conclude that had there been a servo-controller based hydraulic pumping unit or a similar mechanism to maintain the pressure applied on the plate constant, the observed settlements under each loading stage would have been still higher, than that were observed in the field study. Based on the above results, a set of resistance factors is derived for bearing pressures, as shown in Table 2, where resistance factor is defined as the ratio of observed load for maintained case to that of non-maintained case, for an assumed settlement level. The resistance factor varies from 0.63 to 0.67, with an average of 0.65, for settlements in the

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range of 4 to 12 mm. These resistance factors, when multiplied with the loads of non-maintained load test corresponding to any given settlement, can partially compensate the gross overestimation of loads obtained from non-maintained load tests. However, these factors should be used with caution, as they are derived based on limited test data. Similar studies are also conducted in cohesionless soils, and the same trend is observed.

Table 2. Resistance factors for bearing pressures Allowable settlement (mm) 4 8 12 Resistance Factor for load 0.63 0.66 0.67

In conclusion, due care should be taken by the contracting agencies while following the specifications mentioned in the Codes of Practice. It is also mandatory that the local codes should be reviewed in view of the above mentioned uncertainties, and revised from time to time for the benefit of the geotechnical community. 6. CONCLUSIONS Plate load test, which is revered as most reliable in-situ testing technique to obtain load-settlement response, may give rise to unrealistic results if not properly conducted. This paper focuses on the analysis of laboratory plate load test results, to understand the effect of maintained and non-maintained load during each loading stage. Following are the salient conclusions drawn from the study.

1. From both stress maintained and stress not-maintained laboratory plate load tests on remoulded marine clay, it is observed that the load-settlement response for both the cases are different.

2. Load tests with conventional hydraulic pumping units, which cannot hold the pressure constant during loading stages, lead to flatter load-settlement response.

3. Had there been a servo-controller based hydraulic pumping unit or a similar mechanism to maintain the pressure on the plate during each loading cycle, the observed settlements under each loading stage would have been still higher, than that were observed in the present study.

4. The resistance factor for bearing pressure varies from 0.63 to 0.67, with an average value of 0.65, for settlements in the range of 4 to 12 mm. These resistance factors, when multiplied with the loads of non-maintained load test for any given settlement, can partially compensate the gross overestimation of loads obtained from non-maintained load tests.

REFERENCES ASTM D 1194-94, Standard Test Method for Bearing Capacity of Soil for Static Load and

Spread Footings, Annual Book of ASTM Standards, ASTM International, West Conshohocken, PA., 2003.

Briaud, J-L and Gibbens, R. (1999). Behaviour of five large spread footings in sand, Journal of Geotechnical and Geoenvironmental Engineering, ASCE, 125(9), 787-796.

Coduto, D.P. (1994). Foundation Design: Principles and Practices. Second Edition, Prentice-Hall Inc.

Dasaka, S.M., and Jain, A. (2012). Uncertainty analysis of bearing capacity of soils from field plate load tests. Asian-Pacific Symposium on Structural Reliability and its Applications (APSSRA-2012), Singapore, CD-ROM proceedings.


IS: 1888-1982: Indian Standard Practice on Method of Load Test on soils, Bureau of Indian Standards, 2003.

Jain, A., Effect of Loading Mechanism on Plate Load Test Results, M.Tech. thesis, IIT Bombay, 2011.

Lacasse, S. (2001). Karl Terzaghi Lecture.

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99

Prediction of Soil Behavior – A Reappraisal

Binu Sharma

Department of Civil Engineering, Assam Engineering College, Guwahati email: [email protected]

ABSTRACT: Undrained shear strength, consolidation pressure and coefficient of permeability at any water content of remoulded clays are amenable to correlation with the liquid limit and the plastic limit provided the shear strength, consolidation pressures and coefficient of permeability at these limits can be correlated with each other. Based on the premise that soil assumes a unique state at the liquid limit yielding a unique shear strength, unique effective isotropic consolidation pressure and coefficient of permeability, the objective of this write up is to show experimental verification whether the shear strength, effective isotropic consolidation pressure and coefficient of permeability at liquid limit and plastic limit can be correlated with each other. This will result in the formulation of expressions for predicting undrained shear strength, consolidation pressure and permeability of a remoulded soil at any water content based solely on its liquid limit and plastic limit. In this write up expressions for predicting undrained shear strength and consolidation pressure and permeability are discussed. Keywords: plastic limit, liquid limit, undrained shear strength, compressibility and permeability

1. INTRODUCTION In soil engineering, for fine grained saturated soils without any prestressed effects and cementation bonds, the strengths at liquid limit water contents are considered to be unique, with a mean value of 1.7kN/m2. Again according to Russel and Mickle (1970), and Whyte (1982), under an isotropic effective consolidation pressure of about 6kPa all soils equilibriate at their liquid limit water content. Another interesting finding is that the coefficient of permeability k is found to be nearly of the same order at their liquid limit in spite of a wide variation in the liquid limit water contents. Nagaraj et al. (1994), attributes these to the constancy of the physico – chemical potential per unit volume at liquid limit for all soils. Undrained shear strength, consolidation pressure and coefficient of permeability at any water content of remolded clays are amenable to correlation with the liquid limit and the plastic limit provided the shear strength, consolidation pressures and coefficient of permeability at these limits can be correlated with each other. Based on the premise that soil assumes a unique state at the liquid limit yielding a unique shear strength, unique effective isotropic consolidation pressure and coefficient of permeability, the objective of this write up is to show experimental verification whether the shear strength, effective isotropic consolidation pressure and coefficient of permeability at liquid limit and plastic limit can be correlated with each other. This will result in the formulation of expressions for predicting undrained shear strength, consolidation pressure and permeability of a remolded soil at any water content based solely on its liquid limit and plastic limit. In this write up expressions for predicting undrained shear strength and consolidation pressure and permeability are discussed.

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2. UNDRAINED SHEAR STRENGTH As early as 1939, Casagrande suggested an average shear strength of soil at the liquid limit as 2.65 kN/m2 taking into account a large spread of values depending on the apparatus used for determining the liquid limit. Norman (1958) reported that the shear strength at the liquid limit determined by using an apparatus conforming to the British standard ranged from 0.8 to 1.6 kN/m2 whereas using an apparatus of ASTM standards, the strength ranged from 1.1 to 2.3 kN/m2. Skempton and Northey (1953) reported the value of shear strength at the liquid limit of four soils with very different values of plasticity index as 0.7 kN/m2 to 1.75 kN/m2. Youssef et al. (1965) found that the values of shear strength of clay at the liquid limit of a large number of soils (liquid limit varying from 32 to 190%) ranged from 2.4 to 1.3 kN/m2 with a mean value of 1.7 kN/m2. According to Federico (1983), the shear strength at the liquid limit of soils, falls within limits of 1.7 and 2.8 kN/m2. D.M. Wood (1985) showed a mean value of shear strength at the liquid limit as 1.7 kN/m2. Other studies (Russel and Mickel 1970; Wroth and Wood 1978; Whyte 1982; Nagaraj et al. 1990) have shown that shearing strength of all fine grained soils at the liquid limit falls within a narrow range of about 1.7–2.0 Kpa. A detailed explanation for this concentration of values has been presented by Mitchell (1993). According to Wroth (1979), a substantial part of the strength variation at the liquid limit obtained by using the Casagrande apparatus can be attributed to the fact that soil deformation in it is self weight-induced. In the cone test the soil deformation is caused by the cone weight and is essentially independent of the soil weight and hence of its water content. Therefore, it would be expected there would be practically no variation in strength at the liquid limit (in the cone test) with water content. Wroth and Wood (1978) adopted a mean value of 1.7 kN/m2 as the best estimate of undrained shear strength of a remolded soil at its liquid limit.. From the above review a mean value of 1.7 kN/m2 can be adopted as the unique undrained shear strength of a soil at its liquid limit. From the results of four soil samples of Skempton and Northey (1953), Wroth and Wood (1978) assumed that the shear strength at the plastic limit is one hundred times the shear strength at the liquid limit. Based on this assumption Nagaraj et al. (1994) and Belvisco et al. (1985) and many more defined plasticity index as the range of water content producing a 100-fold variation in shear strength. This 100-fold variation of shear strength that is based only on the results obtained by Skempton is verified with test results of 55 inorganic soil samples. This has led to the redefinition of the plastic limit as the water content at which undrained shear strength is around 170 kN/m2. This resulted in the formulation of an expression for predicting undrained shear strength of a remolded soil at any water content based solely on its liquid limit and plastic limit. (Sharma and Bora,2003) A single consistent method (the Swedish fall cone method) was used for determining undrained shear strength of clays from very soft consistencies around the liquid limit to stiff consistencies around the plastic limit. Otherwise disparate methods are in use such as the laboratory vane shear test for clays having shear strength less than 49 kN/m2 and the triaxial compression test for shear strength more than 29.4 kN/m2. Since the liquid limits and the plastic limits are interpreted as particular strength states, it is logical to use a simple test like the fall cone test to measure the strength more directly. Cones of 60gm-60° cone angle, 100gm-30° cone angle and 400gm-30° cone angle have been used. The 60gm-60° cone was used for the investigation of soft clays and the 100gm-30° cone and 400gm-30° cone were used for the investigation of medium and stiff clays respectively. Hansbo (1957) presented a thorough study of the relationship between cone penetration (h) and undrained shear strength (τ) for different cone angles and weights. The relationship is

Prediction of Soil Behavior – A Reappraisal 101

τ = KQ/h2 (1) where Q is the weight of the cone; h = depth of penetration of the cone into the soil; and K = constant the value of which depends solely on cone angle ß for remolded soil. The writer gave the value of the constant K as 0.3 for remolded soil only for the 60gm-60° cone and 10gm - 60° cone. Value of the constant K for the 300 cone angle was not given. 30° cones were calibrated by carrying out unconfined compression test. The K value was found to be 0.82 for these cones after calibration. The liquid limit of the soils for this study was determined by the 60gm-60° cone. Taking undrained shear strength at the liquid limit as 1.7 kN/m2 and by using the above cone, one gets from Eq. (1) 1.7 = KQ/h2 where K = 0.3 or h = 10.2 mm. Thus, water content corresponding to 10.2 mm penetration of the above cone gave the liquid limit of the soil. The undrained shear strengths were determined for 55 inorganic soil samples by the fall cone test, the liquid limit of which varied from 33.8 to 82%. When water content was plotted against shear strength on a log-log scale a straight line for each soil sample was obtained. Fig. 1 shows typical results for the 55 samples tested. The relationship was obtained for water contents greater than the liquid limit to less than the plastic limit of the soil. The linearity in behavior between log τ and log w was also observed for data of three soils taken from the work of Skempton and Northey(1953) who plotted the data originally as liquidity index against shear strength. The result is shown in Fig. 2.

Fig. 1. Relationship between undrained shear strength results as obtained from unconfined compression test and fall cone test and water content

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Fig. 2. Relationship between water content and undrained shear strength (data from Skempton and Northey 1953)

Undrained shear strength behavior was also obtained for water contents much greater than the liquid limit of the soils in order to provide some insight into the strength of ‘‘weak’’ cohesive soils (for example, newly deposited coastal sediments). The results are incorporated in Fig. 1 where it is observed that the linear relationship extends well beyond the liquid limit of the soil. Figs. 1 and 2 also show that the plastic limits as determined by the thread rolling method lie close to a line corresponding to an undrained shear strength of 170 kN/m2. Similarly, the straight line corresponding to an undrained shear strength of 1.7 kN/m2 is the liquid limit line. This confirms that plasticity index can be redefined as a range of water contents producing 100-fold variation in shear strength. Undrained shear strength results at different water contents obtained from unconfined compression tests are also incorporated in Fig. 1. Figure 1 shows good agreement of the cone penetration test results with those of the unconfined compression test. This example is typical of all such results obtained. Using the linear relationship between log τ and log w which is unique for a particular soil and taking undrained shear strength at the liquid limit as 1.7 kN/m2 and at the plastic limit as 170 kN/m2 the following expression can be obtained:

(2)

where τ = undrained shear strength at water content w; τLL = undrained shear strength at the liquid limit; wL = liquid limit; and wP = plastic limit. The remolded strength can be estimated for any value of the water content of a soil by this relationship knowing only the liquid limit and the plastic limit of the soil.


The liquid limit of the 55 soil samples tested ranged from 33.8 to 82%. To examine whether the same relationship is observed in the case of soils having much higher liquid limits, six bentonite samples with liquid limits ranging from 210 to 460% were tested. The liquid limit values of the samples were determined by the 60gm - 60° cone. Undrained shear strengths at different water contents were determined by the fall cone method. A log-log plot of water content versus undrained shear strength yielded a bilinear relationship. Fig. 3 shows two typical results. The first straight line at a lower-water content has a steeper slope. As the water content increases, a second straight line of a lesser slope intersecting the first straight line was obtained. It is also observed that the thread rolling plastic limit values lie around the line having an undrained shear strength of 170 kN/m2. The liquid limit values of three bentonite samples tested in the Casagrande apparatus (shown in Fig. 3) were found to lie around the line having an undrained shear strength of 1.7 kN/m2. This further substantiates the above finding that the undrained shear strength at the plastic limit is 100 times the undrained shear strength at liquid limit. However, due to the bilinear relationship of the above mentioned plot the correlation shown by Eq. (2) breaks down in the case of soils having very high liquid limits. The practical importance of soils having such high-liquid limits is of course comparatively much less.

Fig. 3. Relationship between undrained shear strength and water content of bentonite samples

3. DETERMINATION OF PLASTIC LIMIT OF SOIL BY FALL CONE METHOD The above findings show that soil assumes a unique undrained shear strength at plastic limit also. Taking undrained shear at plastic limit as hundred times the undrained shear at liquid i.e. 170kN/m2, the 400gm - 300 cone had been used to determine the plastic of the soil. The Casagrande method for determination of the plastic limit which is the most commonly used method is prone to human error (Ballard and Weeks 1963). Furthermore, the test

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conditions are more similar to strain-controlled ones than stress-controlled ones (Kenny 1963). There have been many attempts at rationalizing the procedure for determining the plastic limit by using the cone penetration method (Wroth and Wood 1978; Belvisco et al. 1985). All these methods were based on indirect determination of /the plastic limit as it is difficult to devise a penetration test that would perform with equal precision over a 100-fold variation in shear strength from the liquid limit to the plastic limit. According to Hansbo (1957), the penetration h for a 100gm -30° cone is doubled by using a 400gm -30° cone and the latter is thus recommended for the investigation of stiff clays. From the findings of the present work that undrained shear strength at plastic limit can be taken as 100 times that at liquid limit, i.e., 170 kN/m2, and by using the above cone, one gets from Eq. (1) using K = 0.82 170 = KQ/h2 or h = 4.4 mm This shows that the plastic limit can be directly obtained by the cone penetration method corresponding to a specific penetration of a suitable cone. In order to verify the reliability of this procedure, the plastic limit values of 55 inorganic soils were determined by the conventional thread rolling method. Additionally, the 400gm -30° cone was used to determine the plastic limit of the soils as already explained. Fig. 4 shows the result of the conventional thread rolling procedure in comparison to those of the cone penetration method. The results of plastic limit values of the bentonite soils are also incorporated in the same figure. The figure shows that the assumption that shear strength at the plastic limit is 100 times that at the liquid limit, produced values of plastic limit that are in good agreement with the conventional Atterberg plastic limit values. (Sharma and Bora, 2004).

Fig. 4. Comparison of plastic limit values obtained from cone penetration and thread rolling

method

4. COMPRESSIBILITY CHARACTERISTICS It has been shown ( Sharma and Bora 2003, 2004 ) that liquid limit can be determined by the 60gm - 600 cone and plastic limit of soils can be determined by the 400gm 300cone. With


liquid limit and plastic limit so determined a correlation between compressibility characteristic and liquid limit and plastic limit of soil is proposed based on a pressure – void ratio equation derived in case of reconstituted saturated fine-grained soils making use of the experimentally verified fact that the major principal ( vertical ) effective consolidation stresses at liquid limit and plastic limit are around 6.3 kN/m2 and 630 kN/m2 respectively. Wroth (1979) derived theoretically through critical state soil mechanics that at liquid limit the range of the major principal effective stress comes out to be a constant at around 6.3 kN/m2 for a remoulded soil.. (Russel and Mickle, 1970; Whyte,1982 also states that at an isotropic consolidation pressure of around 6kPa, all soils equilibriate at their liquid limit. The experimental verification of this fact was made and it was also investigated to determine the range of the major principal effective stress at the plastic limit of the soil. For this one dimensional consolidation tests were carried out on seventeen inorganic fine - grained saturated soil samples compressing each sample from their slurry state in the fixed ring consolidation cell. The liquid limits of the samples ranged from 33.8 percent to 82 percent with values in the interim almost uniformly distributed. Wroth (1979) showed that when the compression curves of soils are replotted in terms of liquidity index against log of effective consolidation pressure all remoulded soils have one unique normal consolidation line. This can also be applied to a soil exhibiting no sensitivity. Fig. 5 shows the plot of liquidity index (L.I.) against log of effective consolidation pressure (P ) of the seventeen soil samples tested. It shows that when water content is normalised as liquidity index, all remoulded soil test results plot in a narrow band which can be fitted with a linear equation of the form L.I. = 1.42 – 0.506 log P (3) with a correlation coefficient of 0.97. The line generated by Equation 3 shows that at liquid limit (when liquidity index is unity ) the consolidation pressure is 6.3 kN/m2. Similarly, at plastic limit ( liquidity index zero), the consolidation pressure is 630 kN/m2.

Fig. 5. – Liquidity index against log P relationship.

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Nagaraj et al (1991) used the void ratio at liquid limit (eL) as a parameter for generalising the compressibility behaviour of saturated fine-grained soils. The logic for this generalisation was derived from the Gouy – Chapman diffuse double layer theory. To plot the data of the present work in conformity with the method adopted by Nagaraj et al (1991) the e-log P plots of the seventeen soil samples were normalised with their respective void ratios at liquid limit eL. Fig. 6 shows the e/eL against log P plot. The equation of the narrow band can be fitted with a linear equation of the form e/eL = 1.2315 – 0.2933 log10 P (4) with a correlation coefficient of 0.98. Fig 6 shows that at liquid limit the consolidation pressure is around 6.3 kN/m2.

Fig. 6. e/eL against log P relationship. Burland (1990) characterised the intrinsic compressibility behaviour of normally consolidated soil in terms of a new soil parameter, the intrinsic void index Iv , such that

*100

*

1000*

100*

100*

CV C

eeee

eeI −=

−−

= (5)

where e*100 and e*1000 are the intrinsic void ratios at consolidation pressures of 100 kPa and 1000 kPa respectively. The intrinsic compression index Cc* is defined as e*100 - e*1000. The term intrinsic is used to describe the properties of clays which have been reconstituted at a water content between wL and 1.5 wL where wL is the liquid limit water content. The soils are prepared without air drying or oven drying and consolidated preferably under one dimensional consolidation. The chemistry of the water used should preferably be similar to that of the pore water in the clay in the natural state. The term intrinsic has been chosen since it refers to the


basic, or inherent properties of a given soil prepared in a specified manner. Burland (1990) showed that when the intrinsic compression curves of soils covering a wide range of liquid limits are replotted in terms of the void index Iv against log of effective consolidation pressure, a unique line is obtained which is termed the intrinsic compression line ( I C L ). The data of the present study have also been plotted in conformity with the method devised by Burland (1990). Fig. 7 shows the plot of void index IV against log (P ). It is apparent that the ICL of the present study coincides with the ICL obtained by Burland (1990) for most part barring some differences at low magnitudes of stress.

Fig.7. Void index against log P relationship. The average void index at liquid limit [ IvL (ave) ] of the seventeen soil samples tested was found to be 1.19 and that at plastic limit [IvP (ave) ] was found to be – 0.827. Insertion of these values in Fig.7, shows that the consolidation pressure corresponding to IvL (ave)) is around 6.3 kN/m2 and the pressure corresponding to Ivp(ave) is around 630 kN/m2. This lends experimental verification to Wroth’s (1979) theoretical derivation of major principal effective stress of remoulded soils at liquid limits. Major principal effective stress at plastic limit is found to be around 630kN/m2 i.e. hundred times. The void ratio (e) against log of pressure (log P ) relationship of the soils tested has been found to be approximately linear. Using this linear relationship and taking effective consolidation pressures at liquid limit and plastic limit as 6.3 kN/m2 and 630 kN/m2

respectively, the following correlation between void ratio (e) at any effective consolidation pressure P and void ratios at liquid limit and plastic limit of the soil can be obtained:

)/630(log]2/)[( 10 Peeee PLP −+= (6) where e is the void ratio at consolidation pressure P eP = void ratio at plastic limit

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eL = void ratio at liquid limit. The practical advantage of using Equation 6 is that the e – log P curve for reconstituted soils can be predicted with reasonable accuracy without having to conduct any consolidation test. Fig. 8 which is typical of the response of all soils tested under consolidation pressure, shows that there is good agreement between the predicted and the experimental values.

Fig. 8. Experimental and predicted compression curve of three of the seventeen

soil samples tested.

5. PERMEABILITY BEHAVIOUR Griffiths and Joshi (1989) presented pore size distribution patterns for four soils at liquid limit water content, which show the same range of pore diameter, though the liquid limit values vary from 29% to 100%. Pandian, Nagaraj and Sivakumar Babu (1993) stated that the observations of Griffiths and Joshi(1989) suggested that the flow rate at liquid limit should be of the same range/ order. The authors reported coefficient of permeability values at liquid limit ranging from 1.65 × 10-7cm/sec to 2.38 × 10-7cm/sec. Nagaraj et al (1991, 1993,1994) presented the coefficients of permeability values of various clays at liquid limit which ranged from 1.28 × 10-7cm/sec to 3.4 × 10-7 cm/sec. Observing that the coefficients of permeability of different clays are very nearly the same (1.28 × 10-7cm/sec to 2.83 × 10-7 cm/sec ) at liquid limit although liquid limit water contents and void ratios vary over a wide range, Mitchell (1992) concluded that the effective pore sizes controlling fluid flow must be about the same for all clays.


Hence it is observed that the values of permeability coefficients is of the same range /order at liquid limit. As stated before the objective is to verify the above findings and to see whether the same range/ order of permeability occurs at plastic limit also and whether a fixed magnitude of variation occurs in permeability between plastic limit and liquid limit as in the case of undrained shear strength and effective consolidation pressure. Slurry consolidation tests were done to determine the coefficients of permeability at liquid limit. The soils were reconstituted at a water contents slightly greater than their liquid limit water contents and kept for a minimum period of 24 hours in the form of slurry for uniform distribution of moisture. The slurry was then transferred to the oedometer rings. Consolidation tests were carried out using a loading sequence of 5 kN/m2 , 10 kN/m2, 20 kN/m2, 40 kN/m2, 80 kN/m2, 160 kN/m2, 320 kN/m2 and 640 kN/m2. At each pressure after equilibrium was achieved, falling head permeability tests were performed to determine the coefficient of permeability of soil. A thin layer of kerosene was placed over the water to prevent evaporation from the burette. The coefficients of permeability at liquid limit void ratio (eL) were obtained from the plots of void ratio (e) versus log k for each soil. Table 1 summarizes the Atterberg limits, the specific gravity values and the permeability at liquid limit state. The liquid limits of the soils were determined using the 60 gm 600 cone and plastic limits were determined using the 400 gm 300 cone (Sharma and Bora, 2003, 2004). It is observed from the table that even though the liquid limit varies from 33.8% to 78%, the coefficient of permeability at liquid limit is of the same order ranging from 1.28 × 10-7 cm/sec to 3.2 × 10-7 cm/sec. (Sharma and Bora , 2009) This range is consistent with the range given by Nagaraj et al, (1991, 1993, 1994) and by Mitchell (1992) also.

Table 1. Values of permeability at liquid limit water content

SL.No. Liquid limit wL(%)

Plastic limit wP (%)

Specific Gravity

GS

Permeability at Liquid limit

(cm/sec)

Permeability at Plastic limit

(cm/sec) 1 77 28 2.68 2.5 × 10-7 7.5 × 10-9 2 38.5 17 2.63 1.42 × 10-7 8 × 10-9 3 42 20 2.65 1.28 × 10-7 5.9 × 10-9 4 60 24.2 2.71 1.42 × 10-7 7 × 10-9 5 52.5 21 2.71 2.6 × 10-7 8 × 10-9 6 33.8 16 2.68 1.95 × 10-7 9 × 10-9 7 76 29.5 2.74 3.2 × 10-7 9.75 × 10-9 8 45.8 16.2 2.65 1.62 × 10-7 6.5 × 10-9 9 44 16 2.7 2.2 × 10-7 5 × 10-9 10 61 22.5 2.68 2.6 × 10-7 5.7 × 10-9 11 78 29.5 2.71 3.1 × 10-7 9 × 10-9 12 69 24 2.72 2.38 × 10-7 5.9 × 10-9 13 56 24.2 2.694 1.62 × 10-7 7 × 10-9

Coefficient of permeability was also determined at plastic limit which was found to vary from 5 × 10-9 cm/sec to 9.75 × 10-9 cm/sec (Table 1) giving a variation of 95%. In contrast the variation in coefficient of permeability at liquid limit in the same table was found to be 150%. This shows that the variation in coefficient of permeability at plastic limit is even less than that at liquid limit implying thereby that if coefficient of permeability is of the same order at liquid limit so it is at plastic limit. At plastic limit coefficient of permeability which is of the same order (average values 7.25 × 10-9 cm/sec ) is about thirty times less than at liquid limit (average values 2.15 × 10-7 cm/sec ) from Table 1.

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Using the average values of soil permeability at liquid limit and at plastic limit and using the log- linear relationship between void ratio ( e ) and coefficient of permeability ( k) a correlation can be developed between permeability and liquidity index of soils as follows: (7)

or Log ( k) = - 8.139662 + 1.4771213 × L.I. The above equation shows that coefficient of permeability is a log linear function of the liquidity index of the soil. Of course more investigation is needed in this direction. 6. CONCLUSIONS This study emphasizes the fact that there is a certain pattern of variation of soil properties between liquid and plastic limit. Undrained shear strength at plastic limit (170kN/m2) is around hundred times the undrained shear at liquid limit. At plastic limit a soil is subjected to a consolidation pressure of around 630kN/m2 which is again around hundred times that at the liquid limit. It has been verified that coefficient of permeability is of the same range/order ( average value 2.15 × 10-7 cm/sec ) at liquid limit and it has been found that coefficient of permeability is also of the same range or order at plastic limit of the soil ( average value 7.25 × 10-9 cm/sec ). These findings has resulted in the formulation of expressions for predicting undrained shear strength, consolidation pressure and permeability of a remolded soil at any water content based solely on its liquid limit and plastic limit. REFERENCES Ballard, G. E. H., and Weeks, W. F. (1963). ‘‘Human error in determining plastic limit of

cohesive soils.’’ Mater. Res. Stand., 3(9), 726–729. Belvisco, R., Clampoli, S., Cotecchia, V., and Federico, A. (1985). ‘‘Use of cone

penetrometer to determine consistency limits.’’ Ground Eng., 18(5), 21–22. Burland, J.B., 1990 “On the compressibility and shear strength of natural clays.” 30th Rankine

lecture, Geotechnique, 40(3) , pp 329-378. Federico, A. (1983). ‘‘Relationships (Cu-w) and (Cu-s) for remolded clayey soils at high

water content.’’ Riv. Ital. Geotec., XVII(1), 38–41. Griffiths, F.J. and Joshi, R.C., (1989). “Change in pore size distribution due to consolidation

of clays”. Geotechnique, 3991) : 159 – 167. Hansbo, S. (1957). ‘‘A new approach to the determination of the shear strength of clay by the

fall cone test.’’ Swedish Geotech Institute Proc., Stockholm, 14, 1–48. Mitchell, J. K. (1993). Fundamentals of soil behaviour, Wiley, New York. Nagaraj, T. S., Srinivasa Murthy, B. R., and Vatsala, A.(1990). ‘‘Prediction of soil behaviour.

Development of a generalised approach.’’ Ind Geotech. J., 20(4), 288–306. Nagaraj, T.S, Srinivasa Murthy, B.R and Vatsala A. 1991. “Prediction of soil behaviour

.Part 11- Saturated uncemented soils.” Indian Geotech. J., 21:1, pp 137 – 160. Nagaraj,T.S., Pandian,N.S., and Narasimha Raju, P.S.R. 1991. “An approach for prediction of

compressibility and permeability behaviour of sand – bentonite mixes”. Indian Geotechnical Journal. Vol.21, No.3, pp 271- 282.

Nagaraj,T.S., Pandian,N.S. and Narasimha Raju, P.S.R. 1993. “Stress – state – permeability relationship for fined grained soils”. Geotechnique, 43(2): pp 333 – 336.


Nagaraj,T.S., Pandian,N.S., and Narasimha Raju, P.S.R. 1994. “Stress – state – permeability relationship for overconsolidated soil”. Geotechnique, 44(2): pp 349 – 352.

Nagaraj, T.S., Srinivasa Murthy, B.R., and Vatsala A. 1991. “Prediction of soil behaviour. Part 1– Saturated uncemented soils”. Indian Geotechnical Journal, 21:1pp 137 – 160.

Nagaraj, T.S., Srinivasa Murthy, B.R., and Vatsala A. 1994. “Analysis and Prediction of Soil Behaviour”. Wiley Eastern Limited, India.

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Pandian, N.S., Nagaraj, T.S., and Sivakumar Babu, G.L. 1993. “Behaviour of tropical clays. Part 2 – Engineering Behaviour”. Journal of Geotechnical Engineering, American Society of Civil Engineers. 119(5) : pp 840-861

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113

Road Embankments in Water Logged and Frost Affected Areas

– Problems and Solutions

Jai Bhagwan and Kanwar Singh Central Road Research Institute, New Delhi, email: [email protected]

ABSTRACT: The paper presents experiences concerning a few of the many-sided aspects of the problems encountered by CRRI during its participation on earthwork projects all over India. Case studies concerning the behaviour of road embankments, one in waterlogged areas in plain terrain and the other in frost – susceptible areas in Himalayas are described in the paper and the lessons drawn there from are summarized. Keywords: roads, embankments, water logged areas, frost affected areas 1. INTRODUCTION Climatic and other ambient conditions have strong influence on the behavior, response and durability of road embankments, which in turn influence the performance of pavements, built on them. Highway embankments in high rainfall areas and locations where water logging prevails, experience distress from prolonged submergence, seepage and erosion. Yet another important climatic condition that has significant influence on load carrying capacity of fills and sub grades is the freeze-thaw phenomena. Though freezing temperatures are not common in India, yet such conditions prevail at many locations in Himalayas and the various engineering problems arising from freezing temperatures have to be overcome. The paper presents experiences concerning a few of the many-sided aspects of the problems encountered by CRRI during its participation on earthwork projects all over India. Case studies concerning the behaviour of road embankments, one in waterlogged areas in plain terrain and the other in frost – susceptible areas in Himalayas are described in the paper and the lessons drawn there from are summarized. 2. WATER – LOGGED ROAD STRETCH Part of the Udharband – Mahur Road (U M Road) is located in the plains of Cachar valley on the left bank of the river Jatinga. The terrain is flat, with paddy fields. The rains in the region normally extend from April to October and annual rainfall is about 6000mm. The area remains waterlogged for about two to three months of the year during monsoon season in the stretch between Km.21.100 to Km.23.120, whereas, the road stretch between km.19.100 to km.20.700 remains submerged for about one month or more during monsoon season because of back flow of water from river Barak. Water table in the area was found to be located at 1.7 to 2.0m below the ground level during the month of November. 2.1 Soil Type and Pavement Conditions Soil was classified as CI and CL group and PI varies from 10 to 18%. Proctor density varies from 1.58 to 1.65 g/cc indicating that the soil can be used as fill material. CBR of saturated soil ranges 3 to 7%.

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The height of the existing embankment varied from 0.5 to 2.5m for submerged area and from 0.0 to 1.0m above the ground level for water – logged area. Substantial height of embankment found to be eroded and the sub – base Layer of Pavement was found to be exposed. The pavement was in a highly distressed state and longitudinal wave formations were observed frequently over the road pavement. The width of the carriage way was 3.5m and thickness of pavement ranges from 150 to 180mm. Predominant cause for such distress appears to be the use of the local plastic clay with PI ranging from 10 to 20 percent as filler in the sub – base layers. Due to the ingress of water into the embankment fill, the subgrade loses its supporting capacity. This results in pavement crust being thrust into the subgrade under the traffic loads. 2.2. Remedial Measures Recommended Based on the results of field and laboratory investigations, the required improvements call for large scale reconstruction involving;

i. Increase in height and width of the embankments to meet the specified standards as per the guidelines of Indian Roads Congress (IRC).

ii. Increase in pavement thickness, as per traffic and design requirements. iii. Provision of adequate cross drainage works.

2.3. Improvements to the Embankment As per IRC: 37-1984, a free board between the bottom of the pavement and the highest flood level should be between 0.6 to 1.0m. Thus, the minimum required height of the embankment was proposed to be as given below;

i. For the submerged stretch, the height of the existing embankment should be raised by 1.25m from the existing road level (finished surface).

ii. For the water – logged stretch, the height of embankment should be raised by 0.75m above the existing road level.

In both the cases, embankment slope should be 45 degree from the horizontal. The width of the embankment should be increased from the existing width of 5.0m to 7.5m as per IRC: 73-1980, prescribed for (ODR). Since the area is water – logged, a sand blanket layer of 200m is thickness, as per IRC: 34 – 1970, was other District Roads proposed at appropriate level to keep the subgrade in a relatively dry state and preserve the strength of the pavement. 2.4. Composition of Pavement The pavement was designed for a ten year life, as stage construction can be adopted beyond ten years. Average soaked CBR value of 4 was obtained in the Laboratory and was adopted for the design. The thickness of pavement worked out to be 460mm, having the following composition:

i. Compacted sand layer - 200mm ii. Soil gravel layer (gravel 100mm down) - 110mm

iii. Base course (2 layers of WBM of IRC grading 3) - 150mm iv. Premix Carpet (with liquid seal coat) - 20mm

An anti stripping agent of approved quality was proposed at a rate of 0.5 to 1.0 percent by weight of the bitumen in the premix carpet. In soil gravel layer it was recommended that the

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PI of the filler material should be less than 6. 2.5. Cross Drainage Works When the embankment is raised, the general road level would be higher and the deck level of bridges and culverts would have to be raised accordingly. Rising of height of the embankment would disturb the existing flow pattern. Hence, it was considered necessary that adequate drainage to equalize the levels of water quickly on either side of the embankment be provided. For this purpose, Hume pipe culverts of 1.0m dia were recommended to be provided at every 50m distance all along the road embankment. 3. FROST AFFECTED STRETCH IN SIKKIM 3.1. General Description and Site Condition The frost affected stretch was situated in Himalayas at an elevation of nearly 4000 m above MSL. The road passes through this stretch on a relatively gentle bench which has a slope ranging from 10 degree to 30 degree. About 100m above the road, the uphill slope becomes steeper, more than 50 degree, and is indicative of high degree of slope degradation caused by avalanches. The slope mass is washed away from the sleeper benches. The gentle bench represents the zone of accumulation of the hill wash. Thus the formation on this zone consists of broken rock fragments and plant remains mixed in soil. The area receives annual rainfall of about 1000mm. Snowfalls are frequent and on the average 4 to 5 snowfalls have been reported during extreme cold weather which prevails from November to March. During this period, the hillside formation remains in a frozen state. Springs were found on the uphill slope of the road stretch. 3.2. Soil Type and Pavement Condition During the course of field investigations, boreholes were made at the site using hand auger. Black silty soil with organic content was observed upto a depth of 2.8m. Beyond this depth, soil tends to become increasingly rocky in nature. The rock type present at this site was found to be biotitic gneiss. The top 0.50m of subsoil consisted of micaceous silty sand mixed with rock pieces, 0.50 to 1.25m consisted of partially decomposed plant roots and remains and thereafter blackish soil was found upto 2.80m. The organic contents of samples collected from depth of 0.50 to 1.25m were found to be of the tune of 86.3%. The field dry densities of these samples varied from 0.22 to 0.30g/cc and the moisture absorption was as high as 600%. In the field, a considerable amount of water from uphill slope was found to be percolating into the layers below the existing pavement. During the winter months, the water in the soil layers upto a depth of about 0.75m below the bottom of the pavement freezes which thaws as spring sets in and temperature increases. The presence of copious moisture from thaw causes a loss of bearing capacity of the subsoil, thereby, rendering the stretch untrafficable. The pavement has been experiencing severe distress and subsidence during the spring thaw, as the melt water does not drain away from the top Soil Layers.

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3.3. Recommended Remedial Measures In order to protect the bearing capacity of the road bed against freeze and thaw, it was considered necessary to ensure that the subsoil layers below the pavement remains well drained or throughly protected from ingress of moisture. The remedial measures are discussed keeping in view the above objective. 3.4. Protective Measures to Prevent Freeze Thaw Damages Good performance of the pavement can be ensured only if the loss of bearing capacity during the thaw season was prevented. Relative immunity from the adverse effects of thaw can be achieved by providing a thick layer of well draining granular fill. The following steps are proposed for the reconstruction of the roadway in this stretch. As discussed earlier, the hillside formation material upto an average depth of 2m was found to consist of partly decomposed plant root and remains or silty soil mixed with decomposed plant remains. This material has poor drainage properties and suffers severe loss of bearing capacity during thaw. It was, therefore, proposed that the existing subsoil upto an average depth of 2m should be excavated and replaced with granular fill material having good permeability characteristics. The proposed range of gradation suitable for this fill material is given in Fig.1. The expected permeability of this material was of the order of 3.0cm/s. Hard, angular and unweathered rock fragments, which are not susceptible to degradation under wheel load impact, as well as, due to freezing and thawing action of water, were recommended to be used in the fill.

Fig. 1: Proposed Gradation of Granular Fill

Fig.2 shows the cross-section of the proposed reconstruction of the roadway. The height of the fill should be built up in layers by rolling. An uncompacted layer of 300mm may be rolled to about 250mmm thickness. A relatively high permeability of 3.0cm/s is stipulated for the fill to quickly drain away water that may infiltrate from the top or the transient pore water pressures that may be generated due to wheel loads in the spring thaw season. This would ensure a better performance of the road. A geomembrane such as PVC sheet may be laid on the excavated surface as shown in Fig.2. This membrane will serve primarily as separation barrier between the formation soil which has undesirable properties and the granular fill. In addition, the entire top width of the fill was recommended to be provided with impervious

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surfacing to minimize infiltration of water from the top during rainfall or under conditions of long term snow accumulation.

Fig. 2: Proposed Cross Section of Roadway

3.5 Surface and Subsurface Drainage At the bottom of the granular fills, cross – drains consisting of 300mm dia. hume pipes, was suggested to be provided at regular interval of 10m and these should drain towards downhill slope. The pipes should have perforations along the top 2/3 of the surfaces. At the junction of the uphill slope and the fill, a gravel filled trench drain was recommended to be provided parallel to the road length which should extend to 0.5m below the level of the cross drain as shown in Fig.2. A geotextile layer was recommended at the junction of the uphill face and the trench drain. The geotextile layer would allow the seepage from uphill to collect into the drain but prevent the chocking of the drain with soil fines that may be carried by the seepage water from uphill slope. Between the drain and main body of the fill, an impervious geomembrane, viz, PVC sheet may be provided to prevent the seepage water from entering the fill area and channelize it into the side drain. It was proposed that the upper part of side trench drain should be lined so that it will not allow water to percolate down. The depth of open portion should be 0.7m below the top of the road level. It was considered necessary to provide open drain in addition to the trench drain to carry away the surface runoff. If the surface runoff is allowed direct access to the gravel filled drain, the pore spaces in gravel drain would choke and drain would be rendered ineffective. The gravel filled portion of the drain would serve to intercept the sub – surface seepage from the uphill slope. 3.6 Thickness of Pavement The existing pavement consists of an unsurfaced crushed stone layer of 200mm thickness. The new pavement will be placed over the granular fill, which will serve as the base course. Considering the CBR value of this granular layer to be 10%, a pavement thickness of 225mm was required as the traffic intensity corresponds to category C of CBR curves. The thickness of the pavement was recommended to be made up of the following layers:

i. Base course layer 150mm ii. Bituminous macadam 75mm

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Base course was proposed to be laid in two layers of 75mm thickness each followed by bituminous macadam with seal coat. 4. CONCLUSIONS Water – logged Road Stretch In order to ensure that the embankment, near Silchar, does not get submerged during the rainy season, it was recommended that the height and top width of the embankment should be increased. The pavement was designed with a sand blanket layer below the sub – base layer to serve as a capillary cut- off layer. Proper cross drainage works were proposed all along the alignment. Frost – Affected Stretch in Sikkim Suitable remedial measures have been suggested on the basis of the studies conducted in this region. Since the problematic road section was less than 0.5km or so, it was recommended that the existing subsoil upto 2.0m depth should be excavated and replaces with wellgraded granular material encased with a PVC sheet. Gravel filled trench drains provided at the junction of gravel fill and uphill slope. Proper cross drainage works were also recommended. ACKNOWLEDGEMENTS The paper is being published with the kind permission of Dr S Gangopadhyay, Director, Central Road Research Institute, New Delhi. REFRENCES Kalra, I.D. (1988), Maintenance Problems on a Road in Cachar District, IRC Journal, Vol. 49-

2, October, 1988, 317-339. MOST (Road Wing), “Specifications for Road and Bridge Works”, (First Revision).


119

The Principles and Application of Geo-Environmental Engineering

Anil Kumar Mishra

Department of Civil Engineering, Indian Institute of Technology Guwahati, Guwahati email: [email protected]

ABSTRACT: The chemical toxic wastes which are being generated due to rapid industrialization and urbanization not only pollute our ecosystem but also significantly affect the human health. Once released to the atmosphere, these wastes significantly contaminate the soil and ground water. In order to solve this problem a new branch of geotechnical engineering, known as geo-environmental engineering, has emerged. One of the major objectives of the geo-environmental engineering is to remediate the contaminated soil and ground water. Various techniques, which are currently being used to treat the contaminated soil and ground water, are discussed in this paper. Keywords: Geoenvironmental Engineering, contamination, source, remediation, techniques

1. INTRODUCTION The rapid increase in global population and industrialization is resulting in the production of the waste containing a lot of toxic chemicals. Once these toxic wastes are released to the atmosphere, it contaminant our environment. These contaminants affect our ecosystem and adversely impact on human health and environment. Traditionally, the main role of a geotechnical engineer has been to design foundation for structures, investigating the subsurface, designing earth work for dams, retaining structures and investigating problems such landslide, subsidence (Sharma & Reddy, 2004; Taylor, 1948; Terzaghi and Peck, 1948, Lambe and Whitman, 1969). However, due to rapid industrialization, today the geotechnical engineers are facing a new problem emerging from the task of the protecting our environment. Recognizing these problems, a new branch of civil engineering, known as geoenvironmental engineering, has emerged in the early 1990. Geoenvironmental engineering constitutes the study of the behaviour of the soil, rock and ground water when they interact with the various contaminants and addresses the problems of hazardous and non-hazardous waste management and contaminated sites (Sharma & Reddy, 2004). 2. SOURCES OF CONTAMINATION From geoenvironmental point of view, various sources of contamination may be divided into following three groups: i) Sources originating on the ground surface, ii) sources originating above water table (vadose zone), iii) Sources originating below water table (saturated zone). i) Sources originating on the ground surface a) Infiltration of contaminated surface water b) Land disposal of solid and liquid waste c) Accidental spill

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d) Fertilizers and pesticides e) Disposal of sewage and water treatment plant sludge ii) Sources originating above water table a) Waste disposal in excavation b) Landfills c) Leakage from underground storage tanks d) Leakage from underground pipe lines e) Septic tanks iii) Sources originating below water table a) Waste disposal in wet excavations b) Deep well injections c) Mines d) Abandoned or improperly constructed wells Contaminated or polluted soil directly affects human health through direct contact with soil or via inhalation of soil contaminants which have vaporized; potentially greater threats are posed by the infiltration of soil contamination into groundwater aquifers used for human consumption, sometimes in areas apparently far removed from any apparent source of above ground contamination. Chronic exposure to chromium, lead and other metals, petroleum, solvents, and many pesticide and herbicide formulations can be carcinogenic, can cause congenital disorders, or can cause other chronic health conditions. Hence, it is quite essential to treat the contaminated soil and ground water from health point of view. The remedial solution to the geoenvironmental problems can be categorized into two groups; i) contaminated site remediation, ii) waste containment (Sharma & Reddy, 2004). 3. CONTAMINATED SITE REMEDIATION As no environmental laws and regulations were exist until early 1970s, the waste chemicals were disposed of without considering its impact on human health and environment. As a result of this many sites gets contaminated by these toxic chemicals. According to an estimate by U.S. Environmental Protection Agency (USEPA) in 1997, more than 200,000 contaminated sites were present in U.S. alone. In addition to the site contamination resulting from disposing of the chemical wastes, spillage of the toxic chemical during handling and transportation can also pollute the soil. Remediation of contaminated soil within the last several years has become a significant area for application of geo-environmental engineering. Excavation of the contaminated soil followed by transporting it to the disposal site and then to dispose it in landfill is a common procedure followed by the engineers. However, prior to disposing off, the soil is required to be treated to decrease the concentration of the contaminant below the level described by various regulators. The various techniques adopted to treat these contaminated soils are; 3.1. Soil Vapor Extraction This technique is used to remove volatile organic contaminant from the contaminated soil. In this technique, a vacuum is applied to the contaminated soil through the extraction wells, which creates a negative pressure gradient that causes the movement of vapors towards the well. The contaminated laden vapors extracted from the wells and then treated above the

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ground using standard air treatment technologies such as carbon filters and combustion. This method is used to remove volatile organic component from soil and is applicable for the soil which is highly permeable.

Figure 1. A typical soil vapor extraction process (USEPA, 1995) 3.2. Soil Washing This technology is used to separate contaminant from excavated soils and to reduce the volume of soil requiring for final treatment or disposal. The contaminants tend to be associated with organic matter and fine soil particles. After washing the soil, the coarse grained soil is separated out from the contaminant laden fine grained soil and washing fluid. The cleaned coarse grained soil is then disposed in the excavation site and the fine fraction and wash fluid is treated further before disposing of in a landfill. Figure 2 depict a typical soil washing process. Following the excavation, the soil is screened to separate coarse debris (larger than about 2 in.) such as rocks. The remaining soil is fluidized with the addition of water. In the scrubbing unit, a water based washing solution is used to separate soluble contaminants and fine particles from coarser soil materials. Surficial contaminant is removed from the coarse fraction by solution. The scrubbing action also disintegrates soil clumps, freeing contaminated fine particles from larger grains. Soil washing can be effective for treating soils contaminated with variety of organic and inorganic contaminants (Semer and Ready, 1996). This process is not so effective for treating silty or clayey soil. 3.3. Stabilization and Solidification (S/S Process) The stabilization and solidification process, also known as immobilization, fixation or encapsulation, uses additives or processes to chemically bind and immobilize contaminants that physically prevent movement of contaminants. Stabilization typically refers to a chemical process that actually converts the contaminants into a less soluble, mobile, or toxic form. Solidification generally refers to a physical process where semi-solid material or sludge is treated to render it more solid.

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Figure 2. Various steps involve in a soil washing process (USEPA, 1996)

This technology neither removes the contaminants from soils, such as soil washing, nor degrades the contaminants, such as bioremediation. It only eliminates the mobility of contaminants. The stabilization and solidification (S/S) process can be implemented ex-situ or in-situ conditions. An ex-situ S/S process involves the following steps; i) Excavation of contaminated soil, ii) Mixing reagent with the soil, iii) Curing of the mixed product, iv) Backfilling or landfilling of the treated soil. An in-situ S/S process involves the injection and/or mixing of stabilizing agents into subsurface soils to immobilize the contaminants, to prevent them leaching into ground water. The S/S process applicable to soil contaminated with metals, radio nuclides, and other inorganic as well as non-volatile and semi-volatile organic compounds. Soil contaminated with volatile organic compounds are not considered to be appropriate for the S/S process because they may be volatilized and released during mixing and curing operations. The S/S process is applicable for all type of soils (clays, silts, sands). 3.4. Electro-kinetic Remediation In this process the contaminants are removed by applying an electric potential. This technology involves applying an electric potential across contaminated soil through a pair of electrode, located at cathode and anode. Then the contaminants are transported towards the electrode and then the contaminants laden liquids are removed from electrode. The system consists of minimum two electrodes buried under ground and connected to a power supply. The electrodes are located certain distance apart and are encased by reservoir or wells. The electrodes are called anode (+ve charged) and cathode (-ve charged). Anode attracts contaminants that have a -ve charge and cathode attracts +ve charge contaminants. In remediating unsaturated soil, water is injected into the ground. Removal of contaminants can be achieved by pumping the contaminated water in the reservoir or wells


Figure 3. A typical electro-kinetic remediation process (USEPA, 1997)

3.5. Bioremediation Bioremediation is a process in which micro-organisms degrade organic contaminants or immobile inorganic contaminants. Under favorable conditions, micro-organism can degrade organic contaminants completely into non-toxic by-product such as carbon dioxide and water or organic acid and methane. In natural attenuation process, micro-organisms occurring in the soil (yeast, fungi, or bacteria) degrade the contaminants for their survival. However, depending upon the type of contaminant and its toxicity levels, specific microbes may be introduced into the soil to be remediated. For microbial growth and survival, supplied of oxygen, moisture, and nutrients may be needed. The process of bioremediation refers to enhancement of the natural process by adding micro-organism to the soil, referred as bio-augmentation, and/or supplying oxygen, moisture, and nutrients required for microbial survival and growth to the soil, referred to as bio-stimulation. Bioremediation process which occurs in the presence of oxygen or air is called as aerobic bioremediation. Under aerobic conditions, contaminants are converted into carbon dioxide and water. Bioremediation process which occurs in the absence of oxygen or air is called anaerobic bioremediation. Under anaerobic condition, organic contaminants are converted into methane, limited amount of Carbon dioxide and traces of hydrogen. Aerobic bioremediation process is quicker compare anaerobic bioremediation. At most sites, micro-organism naturally exists that capable of degrading the contaminants. However, environmental conditions are not conducive for these micro -organisms to degrade the contaminants. Bio-remediation basically involves supplying oxygen, moisture, and nutrients to the contaminated soil zone so that the naturally existing micro-organisms are activated to degrade the contaminants. For degradation to occur, it has to be ensured that oxygen, moisture, and nutrient concentrations are maintained in sufficient amounts. The monitoring can be done by maintaining monitoring wells and also by measuring concentration of carbon dioxide and oxygen. The increase in biological activity will be marked by the decrease in oxygen concentration and an increase in carbon dioxide concentration. Bio-remediation is commonly used for the treatment of soils contaminated with organic compounds. This process cannot degrade inorganic contaminants such as heavy metals. This process can be used for any kind of soil with a sufficient amount of moisture content.

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Figure 4. Degradation of contaminants by microbes (USEPA, 1991)

3.6. Phytoremediation Phytoremediation involves removal, stabilization or degradation of contaminants in soils by plants. The various remediation mechanisms are in the root zone or in the plant itself. Plants are in contact with the contaminated soil in the root zone. Pollutant must pass through the root membranes before they are absorbed by the plant. Contaminants fate is determined by the plant’s capacity to break down the absorbed organic chemicals by plant metabolic process, called phytodegradation, or in-corporate inorganic chemicals in plant tissue, called phytoaccumulation. The above ground portions of the plant are then harvested and burned or decomposed. There are various steps involve in a phytoremediation process. These are;

a) Phytostabilisation: Contaminants are absorbed and immobilize by root along with the ground water.

b) Rhizodegradation: It is the breakdown of the contaminant by microbes present in soil. c) Rhizofiltration: Passing of the contaminants through root membrane before they are

absorbed by plant. d) Phytodegradation: Breaking down of the absorbed contaminant by plant’s metabolic

process. e) Phytovolatilization: Release of the water soluble contaminants to the atmosphere.

Phytoremediation is best suited at sites with shallow (<10 ft) contamination. It is well suited for use at very large field sites and sites with low contaminant concentration. Generally, plants like willow, Poplar, Indian mustard, sunflower, cabbage are used to treat contaminants such as explosive, crude oil or oil products, pesticides, heavy metals, radio nuclides.

3.7. Pump and treat method This is the most common technology used for ground water remediation. The treatment system involves pumping ground water to the surface, removing the contaminants and recharging the treated water into ground or discharging it to the surface water or to municipal sewage plant. When ground water has been pumped to surface, it can be treated to reduce contaminants to a very low level. Pump and treat system can be designed to meet two objectives: i) Containment- To prevent contamination from further spreading; ii) Restoration- To remove contaminant mass.


Figure 5. A schematic diagram of pump and treat method (Ready& Sharma, 2004)

3.8. Permeable Reactive Barrier (PRB) Permeable reactive barrier (PRB) is a ground water treatment method. It is a treatment technology design to degrade or immobilize contaminants contained in ground water as it flows through the barrier. A barrier is put in place by constructing a trench across the flow path of contaminated ground water and filling it with a reactive medium. As contaminated ground water passes through the PRB, the contaminants are either immobilized or transformed into non-toxic substances. Therefore, PRB is not a barrier to the water, but it is a barrier to the contaminant.

Figure 6. Permeable reactive barrier for ground water treatment (USEPA, 1997)

Table 1. Contaminants treated by different reactive media in PRB Contaminants Reactive Media Organic DCE, TCE, PCE DCA, TCA PCB Inorganic Heavy metals (Ni, Pb, Cd, Cr, Hg)

Zero valent iron, Iron (II)-porphyrins Dithionite Zeolite Peat, Ferric oxyhydroxide,

(Ready& Sharma, 2004)

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4. WASTE CONTAINMENT Waste containment is required to prevent further spreading of contaminant when the subsurface soil and groundwater are contaminated with toxic chemicals. Generally, the wastes are contained by providing barriers around it. These barriers are in the form of vertical barrier, bottom barrier, surface cap, and ground freezing technique. 4.1. Vertical Barrier Vertical limits the lateral spreading of the contaminants by redirecting or blocking the ground water flow.

Figure 7. A general configuration of waste containment using vertical and bottom barrier (Ready& Sharma, 2004)

Vertical barriers can be constructed in many configurations. It can be extended up to the bottom barrier, as shown in Fig 7, or it can be embedded up to a low permeable formation. If the contaminants are floating on the ground water table, a hanging type vertical wall extending beneath the water table can be used. The horizontal configuration of a vertical barrier can be circumferential, up-gradient, or down-gradient. With circumferential configuration, the vertical barrier completely surrounds the waste site. This particular layout is preferred if ground water flow conditions are uncertain or unknown. An up-gradient configuration is used to prevent ground water flow through the contaminant and consequent contaminant spreading. A vertical barrier with down gradient configuration, in conjunction with ground water extraction wells, is used to allow ground water flowing through the contaminated zone and then to flush the contaminant from the site.

Figure 8. Horizontal configuration vertical barrier (Ready& Sharma, 2004)

a. Circumferential barrier b. Up-gradient barrier c. Down-gradient barrier


Totally impervious barriers are impossible to construct. However, the barriers can be built to meet the low-hydraulic conductivity requirements of the governing regulatory agencies. They can be made resistant to chemical transport by advection as well as diffusion. The most common type of vertical barriers are, i) compacted clay barrier, ii) Slurry trench barrier, iii) Grouted barrier, and iv) Steel sheet pipe barrier. 4.2. Bottom Barrier Bottom barriers (Fig. 7) are used when no naturally occurring low-hydraulic conductivity stratum exists at reasonable depth beneath a waste site. This construction can be accomplished in several ways, such as by using grouting techniques or employing a combination of tunneling, installation of geomembranes, and grout or slurry mix. 4.3. Surface Cap Surface caps are constructed over buried waste to prevent the infiltration of precipitation, thereby minimizing the generation of leachate. Caps also prevent the transfer of contaminants to the atmosphere. The type of cap depends primarily on the nature of the waste, site condition. It may range from a one layer system of vegetated soil to a multi-layer complex system of soil and geosynthetic.

Figure 9. Surface cap cross-section (Ready& Sharma, 2004)

4.4. Ground Freezing Technique Ground freezing technique can be used to contain polluted ground water. In saturated soils, ground freezing forms a semi-permeable wall as ice crystals form and fill the pore spaces. If the soil is not saturated, then the injection of additional water may be necessary to form ice barrier. The freezing pipes are installed around the contaminant plume to be contained and then the freezing front will spread out from the pipes until a complete wall is formed.

Figure 10. A schematic diagram of ground freezing technique

128 ETGE 2012

5. CONCLUSIONS Various techniques to remediate contaminated ground water and soil were discussed. Each of these methods has their own merits and demerits. Many of these methods are not only time consuming and expensive but also not full proof. Hence, further research should be carried out to find an effective and efficient way to remediate ground water and soil. In addition to this, general awareness should be created among the public for the protection of the environment. REFERENCES Lambe, T.W., AND Whitman, R.V. (1969). Soil Mechanics, Wiley, New York. Semer, R., and Reddy, K.R. (1996). Evaluation of soil washing process to remove mixed

contaminants from sandy loam, Journal of Hazardous Material, 45(1), 45-57. Sharma, H.D., and Reddy, K.R. (2004). Environmental Engineering, Wiley, New York Taylor D.W. (1948) Fundamentals of soil mechanics, Wiley, New York. Terzaghi, K., and Peck, R.B. (1948). Soil Mechanics in Engineering Practice, Wiley, New

York. USEPA (United State Environmental Protection Agency) (1991). Understanding

bioremediation: A guidebook for citizens, EPA/540/2-91/002, Washington D.C. USEPA (United State Environmental Protection Agency) (1995). Soil vapor enhancement

technology resource guide, EPA/542/B-95/003, Washington D.C. USEPA (United State Environmental Protection Agency) (1996). A citizen’s guide to soil

washing, EPA/542/F-96/002, Washington D.C. USEPA (United State Environmental Protection Agency) (1997). Recent developments for in

situ treatment of metal contaminated soils, Washington D.C. USEPA (United State Environmental Protection Agency) (1997). Permeable reactive

subsurface barriers of the inspection and remediation of chlorinated hydrocarbon and chromium (VI) plumes in ground water, EPA 600-F-97-008, Washington D.C.

Proceedings of National Workshop on Emerging Trends in Geotechnical Engineering (ETGE 2012), Guwahati, 08th June 2012 Edited by A. Murali Krishna Department of Civil Engineering Indian Institute of Technology Guwahati Guwahati.

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