EQ Design Actions

7/30/2019 EQ Design Actions

1/34

139

66EARTHQUAKE INDUCED

FOUNDATION ACTIONS

The usual load that a foundation is required to sustain is the vertical loadarising from static permanent and imposed actions. However, horizontal loads

and moments are also common and may be significant. Horizontal loads arerarely applied to a structure exactly at foundation level, thus the horizontal loadis usually accompanied by a moment which simply reflects the distance abovefoundation level at which the horizontal shear is applied.

In static situations a common example of a foundation subject to vertical load,horizontal shear and a corresponding moment is the base of a gravity retaining

wall. The function of the weight of the wall is to transfer a horizontal forcedown to a foundation level where the required reaction can be generated. Asimilar example is the horizontal thrust that is applied to a dam by the waterimpounded behind.

In dynamic situations there are three common sources of foundation loadingwhich apply vertical and horizontal loads as well as overturning moments;these are earthquake, wind and wave loading.

A special sub-class of dynamic loading comes from machine foundations. Thistopic is not given explicit treatment in this book as there it is the steady stateresponse of the system, involving tens of thousands and even millions, ofcycles that is significant. On the other hand, earthquake response can onlyrarely be regarded as steady state.

Regardless of the approach used to arrive at design loads or a design groundmotion the details of the soil profile encountered at the site under


2/34

Design of Earthquake Resistant Foundations

140

consideration are important. First, the soil layer will alter the frequency contentof the earthquake motion transmitted upwards towards the ground surface,thus, for a given earthquake, the response spectra recorded at the top of a soilprofile and at an adjacent rock site will probably have the largest spectral

accelerations at different periods. Second, the presence of the soil layersinfluences the intensity of the recorded ground motion relative to that of anadjacent rock site. The effect depends on the level of seismic excitation: (i) atlow levels of excitation, the soil behaves elastically and acts as a filter and tendsto amplify those components of the motion with periods corresponding to thenatural periods of the soil layers; (ii) at large levels of excitation, the strength ofthe soil provides a limit on the shear stresses that can be transmitted upthrough the soil profile, and so nonlinear stress-strain behaviour of the soilbecomes important.

Additionally, there is the interaction between the building foundation and the

soil supporting it. This is known as Soil-Structure-Interaction (SSI); someintroductory material is given in the next chapter. The interplay between siteeffects and soil structure interaction is illustrated in Figure 6.1. Several of themeasurements of site and foundation response are illustrated in Figures 6.2 to 6.8.

6.1 SITE EFFECTS

First we need a definition of the term site effect, which imagines that we haveadjacent sites, one with rock at the surface, and the other on a soil profileoverlying the rock. As the sites are adjacent we can assume that the incoming

rock motion will be the same, or nearly so, for both. Then any difference betweenthe motions recorded at the two sites can be attributed to the differing groundprofiles, Figure 6.1, and is referred to as a site effect. The velocities andaccelerations at the surface of the soil may be greater or less than those at thesurface of the rock, either is referred to as a site effect. Site effects might also beindicated by differences in frequency content of spectra. Notice that usually wecompare motions at the free surface. However, comparison between the rockmotion at the base of the soil column with the motion at the surface is another

way of indicating the way soil layers modify earthquake motions.

Figure 6.1Site effects and soil-structure interac tion

bedrock motion

(site) amplification of groundmotion by far-field soil body

site-modified spectrafor surface groundmotion

near-field soil modelledby discrete elastic springs

A, B

C

D

E


3/34

Chapter 6: Earthquake induced foundation actions

141

Figure 6.2Aftershock recording, surface geology, and seismograms for

three sites in Oakland (after Borcherdt and Donovan (1989)).


4/34


142

There are various ways in which this difference might be quantified. The simplestis to take the ratio of the peak ground acceleration (PGA) at the surface of the soilprofile to that at the surface of the rock site. The Loma Prieta records in theOakland area provide an example of this type of effect. The PGA at rock sites was

of the order of 0.08g and those at the surface of soil profiles were up to threetimes greater, Figure 6.2. Although there are several kilometres between the sitesof these instruments, all can be regarded as being the same distance from thezone of energy release of the aftershock. It is very clear that the recordedmotion is greatly affected by the type of ground on which the instrument islocated. In some cases an adjacent rock motion was not available but instrumentsare positioned at various depths in the soil profile and possibly in the underlyingrock. The ratio of the PGA at the ground surface to that in the rock underlyingcan be used, as mentioned above, as a measure of the site effect, but it is not thesame as the ratio of the soil and rock site peak ground surface accelerations.

Another possible quantification of a site effect is to take the ratios of the peakspectral accelerations for the rock and soil surface motions.

The above are single figure indications of site effects. Measures over the range offrequencies can be obtained by taking ratios of response spectra, that is at eachperiod the ratio of the spectral acceleration for the soil surface spectrum is dividedby the corresponding value for the response spectrum of the recorded motion atthe rock site. A related method is to take Fourier spectral ratios, either velocitiesor accelerations, rather than ratios of response spectra. Because of the "spikey"nature of many Fourier spectra smoothing is usually necessary. These spectralratio comparisons tend to emphasise the features of the soil profile and smooth

out the particulars of the incoming rock motion.

Decreasing PGA and spectral ratios with increasing input excitation are indicativeof nonlinear soil behaviour.

A review of published data reveals that, at many instrumented sites, the soil profileresponds in an approximately linear manner. If these sites are at the top of acolumn of stiff soil it is not surprising that they behave linearly. Neverthelessmany recent earthquakes, either by direct measurement (Loma Prieta (1989) andMexico City (1985)) or by inference (Philippines (1990), Newcastle (1989),

Armenia (1988)), have confirmed earlier evidence that soil sites will behave

differently from rock sites. For small to modest peak accelerations in theincoming rock motion it is very clear that the soil layers amplify the response. Atlarger accelerations there is the possibility that the soil will deform in a nonlinearmanner; if this happens weak motion data cannot be extrapolated to give strongmotion predictions. The question of nonlinear soil behaviour has beencontroversial but there have been some significant developments in recent years.Some examples are given in the next section.


5/34


143

Figure 6.3 Spectral ratios for the Coalinga strong and weak ground motions

(after Jarpe et al 1988).

6.1.1 Examples of site effects from recorded earthquake motions.

Coalinga, California

Strong motion data was recorded at two sites near Coalinga during the 1983earthquake, in addition aftershocks were recorded. The peak horizontal groundaccelerations recorded ranged between 0.03g and 0.72 g. In addition 23 weakmotion events were recorded in April and May 1985 with two digitalseismometers co-located with the strong motion instruments. There were twoinstrument sites a few kilometres apart. The soil site consists of up to 150m ofalluvium underlain by Cenozoic and Cretaceous sedimentary rocks. The alluviumis said to be unsaturated but no shear wave velocity data are given. The rock site istertiary age sandstone. Figure 6.3 has spectral ratios, obtained from smoothedFourier spectral amplitudes, calculated by Jarpe et al (1988) for the strong and

weak motions. For the 1 to 10Hz frequency band the average strong motionspectral ratios are similar to the average weak motion ratios, although Jarpe et alnoted that scatter for the strong motions is considerably greater than for the weakmotions. Consequently the Coalinga sites seem, on average, to have behavedlinearly.

Mexico City

The Mexico City response to the 1985 earthquake has been thoroughlyinvestigated. As explained by Whitman (1986) there are several features of the siteconditions in Mexico City which are unusual and even unique. One of these is thelarge strain range over which the volcanic clay behaves in an elastic manner,consequently during the 1985, and other, earthquakes the local soils amplified theincoming rock motion elastically, Romo et al (1989). The epicentre of the 1985earthquake was about 360 km from Mexico City consequently the intensity of theincoming rock motion was quite modest, Figure 6.4. However, the overlying soilresponded in an approximately elastic manner so the response spectrum in Figure6.4 shows significant amplification at the elastic period of the soil layer, about 2seconds at the SCT site and about 3.5 seconds at the CAO site. Thesepredominant periods are about the same as the natural period of the 'bowl' of


6/34


144

soft soil beneath the city. It was these large spectral accelerations at the groundsurface that caused the damage to so many of the buildings in the city.

Loma Prieta 1989

The records for the Treasure Island and Yerba Buena Island instrument pair,which are separated by only a few km, have received intense investigation sincethe Loma Prieta earthquake in 1989, Jarpe et al (1989), Idriss (1990), Hryciw et al(1991) and Dickensen et al (1991). The soil profile at Treasure Island consists of13m of hydraulically placed sand, overlying 18 metres of Young Bay Mud,followed by 10m of dense sand and then 45m of Old Bay Sediments to

Figure 6.4 Response spectra from the motions recorded at various locations inMexico City during the 1985 earthquake (after Seed et al 1988).

Figure 6.5 Spectral ratios for the Loma Prieta mainshock and aftershocks at

Treasure Island; solid line: main shock, shaded band: 95% confidence interval

on aftershock data (after Jarpe et al 1989).


7/34


145

Figure 6.6 Observed and predicted alluvial site peak ground acceleration forthe Loma Prieta earthquake (after Chin and Aki 1991).

bedrock. A shear wave velocity profile was obtained using the Seismic ConePenetration Test. The PGA recorded on rock at Yerba Buena Island was 0.06gand 0.16g at the nearby Treasure Island. The ratio between these two is of thesame order as ratios in nearby Oakland but in the case of Treasure Island there isgood evidence of nonlinear soil behaviour. Jarpe et al (1989) and Darragh andShakal (1991) computed the spectral ratios between the two sites for the main

shock and also for aftershock records captured at the same locations. The resultsare shown in Figure 6.5 and provide clear evidence of nonlinear behaviour at the

Treasure Island site. This is of interest because the PGA recorded at TreasureIsland was nearly three times that at Yerba Buena Island, which suggests thatamplification between a soil site and an adjacent rock site is not necessarilyevidence for linearity, unlike the situation in Mexico City in 1985. Liquefaction

was observed to occur at Treasure Island after 15 seconds. The strong motionspectral ratios in Figure 6.5 were calculated from the first 5 seconds following thearrival of the main shock S-wave, so liquefaction will not have affected the ratios.

The contribution of Aki and his co-workers

Several recent papers on site effects have come from Aki: Aki (1988), Aki andIrikura (1991), Chin, Aki and Martin (1991) and Chin and Aki (1991). Thesepapers have moved from a position that seismological evidence reveals a goodcorrelation between weak and strong ground motion, to a statement thatnonlinear soil behaviour may have been more significant than most seismologistshave thought, to the presentation of evidence for nonlinear behaviour. Chin and

Aki (1991) report on the use of weak motion site amplification factors derivedfrom coda waves of the Loma Prieta strong motion records. They then usedthese to predict the PGA for the strong motion event. Figure 6.6 shows how thistechnique overestimates the PGA for distances less than 80 km from the source.Chin and Aki concluded that there exists a pervasive nonlinear site effect in the

epicentral region of the Loma Prieta earthquake.


8/34


146

The Lotung experimental site, Taiwan

At this site two downhole arrays have been installed at various depths in a soilprofile, Chang et al (1989 and 1990). Instruments were installed at the surface andat depths of 6, 11, 17 and 47 metres. At the test site the soil profile consists of

about 50m of recent alluvium underlain by 350m of pleistocene formation. Ashear wave velocity profile is available for the upper 50m or so of the profile. Theupper 30m of the profile consists of silty sand and sandy silt with some gravel.Beneath this the soil is predominantly clayey silt and silty clay. The water table is

within 0.5m of the ground surface. A number of moderate to strong earthquakeshave been recorded at the Lotung site with PGAs in the range 0.03 to 0.26g. Asno nearby rock outcrop record is available Chang et al consider a number of onedimensional analyses to arrive at change from the in situ shear wave velocityrequired to produce the Fourier spectrum of the recorded motion. Theycompared the measured small strain shear wave velocities for the soil profile withthe computed apparent shear wave velocities in the soil profile during the passage

of the earthquake. Evidence for nonlinear

Figure 6.7Effect of peak ground acceleration on site response. Upper: Usual

understanding prior to the mid-1980s, Lower: the understanding after the

1985 Mexico City and 1989 Loma Prieta earthquakes (after Idriss (1990)).


9/34


147

Figure 6.8 Average spectral shapes for three different subsurface conditions(after Idriss (1985)).

behaviour is provided by this reduction of shear wave velocity.

6.1.2 From the above observations to some tentative general conclusions

The recorded site response data in Figure 6.2 are for small levels of groundshaking. As mentioned above the response of a soil site varies as the intensityof the ground shaking increases for very small shaking intensity the soilresponds in an elastic manner and as the intensity of the motion increases thenonlinear stiffness and damping behaviour of the soil come into play, as shownin Figures 3.24 and 4.41. Prior to the 1985 Mexico City and 1989 Loma Prietaearthquakes the geotechnical community expected that nonlinear soil stress-

strain behaviour would reduce the intensity of earthquake shaking at most soilsites as shown in the upper part of Figure 6.7. However, both these


10/34


148

earthquakes showed that in situ cohesive soils are elastic over a great rangeof strains than had been expected. The bottom part of Figure 6.7 presents aninterpretation of these effects from Idriss (1990). This shows that for smallearthquakes soil profiles can indeed be expected to amplify the incoming

earthquake motion. As the level of excitation increases the amplificationdecreases but, from the lower part of Figure 6.7, only quite large peak groundaccelerations reduce the response of the soil site to less than that of the rocksite.

There have been a number of suggestions as to how the shape of the responsespectra vary for different sites, one example from Idriss (1985) is reproduced inFigure 6.8. The second part of this diagram provides a means of scaling thespectrum for different earthquake magnitudes. The spectral shapes given inFigure 6.8 have been normalised by dividing all the ordinates by the peakground acceleration. From this it is apparent that, for normalised spectra

anyway, spectral shapes for different site conditions differ not so much inamplitude but in the period of the maximum response.

6.1.3 Numerical calculations of one-dimensional site response

Further insight into the conclusions of section 6.1.2, and the informationconveyed in Figure 6.2 to 6.8, can be achieved by calculation of the response ofsimple soil profiles to vertically propagating shear waves, both steady statesinusoidal waves and earthquake waves. If the materials behave elastically it ispossible to calculate exactly the response of a layered system to earthquakeexcitation; see Kramer (1996), Towhata (2008), Wolf (1985) and Schnabel et al

(1972). Figure 6.9 shows the results of such calculations when a layer of soil 20m thick overlies and elastic half space. The figure gives the steady stateresponse to a number of sinusoidal waves at differing frequencies, the responseto each frequency is an independent calculation. Figure 6.9 plots theamplification in the surface motion with respect to the input at the base. Itshows that the natural frequency, or first mode (the largest peak), isaccompanied by many other frequencies or modes. Figure 6.9a illustrates theeffect of damping and shows how as the damping increases from 2.5% to 10%the amplification is approximately halved. The amount of amplification is afunction of the contrast in shear wave velocity between the soil and theunderlying rock; for the particular calculations shown in the figure the shear

wave velocity of the rock is 2000 m/sec. and that of the soil is 200 m/sec.Figure 6.9b, for a rock layer with a shear wave velocity of 2000 m/sec. anddamping of 5% illustrates the effect of the shear wave velocity of the soil layer

with a damping value of 2.5%. It is apparent that the amount of theamplification depends on the contrast between the shear wave velocity of thesoil and the rock.

It is possible to extend the above calculations to modelling the earthquakeresponse of a soil layer, provided the soil and the rock respond elastically to theearthquake motion. The results of such calculations are shown in Figure 6.10.

The acceleration response spectrum of the earthquake motion applied at the

base of the soil stratum is compared with that at the top of the soil layer in


11/34


149

0 5 10 15 20 25 300

2

4

6

8

10

Frequency (Hz)

Amplification

8.39

6.30

4.21

0 5 10 15 20 25 300

2

4

6

8

10

Frequency (Hz)

Vs =100 m/sec

Vs =400 m/sec

Vs =1600 m/sec

a b

Figure 6.9 Elastic steady state response to sinusoidal excitation, at variousfrequencies, of a 20 m thick elastic soil layer overlying a deep elastic rocklayer (damping and shear wave velocity of the underlying rock 2.5% and2000 m/sec respectively). (a) Soil damping values of 2.5%, 5.0% and 10%(soil shear wave velocity 200 m/sec), (b) Effect of the shear wave velocityof the soil (soil damping value 5%).

Figure 6.10a. The key to this calculation is Figure 6.10c, which, just like Figure6.9a, shows how the various frequencies are modified by the properties of thesoil layer through the waves propagate. Figure 6.10b has the Fourier spectrumof the earthquake motion applied to the base of the soil layer and Figure 6.10dhas the Fourier spectrum of the motion at the top of the soil layer. One getsfrom Figure 6.10b to Figure 6.10d by taking the product of the Fourierspectrum of the input motion, Figure 6.10b, with the amplification function inFigure 6.10c. This gives the Fourier spectrum of the motion at the top of thesoil layer. The time history of the motion at the ground surface is obtained bytaking the inverse Fourier transform of the spectrum in Figure 6.10d. It is

worth noting in Figures 6.10b and 6.10d how the amplification function inFigure 6.10c generates "humps" in the Fourier spectrum of the motion at theground surface corresponding to the first three humps in the function inFigure 6.10c.

The calculations plotted in Figures 6.9 and 6.10 are for an elastic soil layeroverlying elastic rock. However, we know from Figures 3.24 and 4.41 that soilbehaves nonlinearly with the apparent shear modulus being a function of theshear strain amplitude. Consequently we should not expect a soil deposit torespond elastically to the passage of an earthquake. Software for calculating theelastic response of a layer to vertically propagating earthquake motion isfrequently also able to calculate the nonlinear soil response. One of the earliest

ways of doing this is SHAKE (Schnabel et al 1972), which, using an iterativeprocess, modifies the modulus of the soil at each position in the layer toaccount for the shear strain generated. Known as the equivalent linear methodthis calculation is still an elastic one but the modulus of the soil layers is not thesame as the small strain shear modulus derived from the shear wave velocity.More recently DEEPSOIL, (Hashash et al 2011), has become available


12/34


150

0 0.5 1 1.5 20

0.5

1

1.5

Period (sec).

Spectralacceleration(g).

Ground surface response

Input

a

0 5 10 15 200

0.2

0.4

0.6

0.8

Frequency (Hz)

Amplitude(g-sec)

b

0 5 10 15 200

2

4

6

8

10

Frequency (Hz)

A

mplification

b

0 5 10 15 200

0.2

0.4

0.6

0.8

Frequency (Hz)

Amplitude(g-sec)

d

Figure 6.10 Response of an elastic soil layer overlying a deep layer of elasticrock to an earthquake motion which comes through the rock and into thesoil. (Soil layer 20 m thick, with shear wave velocity 100 m/sec. and

damping of 5%, rock shear wave velocity 2000 m/sec. and damping 5%.)(a) Acceleration response spectra for the rock motion and the groundsurface response. (b) Fourier amplitude spectrum of the rock motion. (c)Amplification function for the elastic soil layer. (d) Fourier amplitudespectrum of the ground surface response.

with the capability of linear, equivalent linear, and several approaches to truelynonlinear modelling of soil stress-strain behaviour. Some of these nonlinearmethods are based on representing data in Figures 3.24 and 4.41 usinghyperbolic relationships like that shown in Figure 3.17. Figure 6.11 presentsthe results of calculating the response of the soil layer / rock layer combination

discussed above to the passage of the same earthquake motion as used in

c


13/34


151

0 1 2 3 40

0.5

1

1.5

2

2.5

Period (sec).

Specctralacceleration(g).

T =0.27 sec. natural period of soil layer

a

0 1 2 3 40

0.5

1

1.5

2

2.5

Period (sec)

T =0.27 sec. natural period of soil layer

b

Figure 6.11 Nonlinear response of a soil layer overlying rock to vertically

propagating earthquake waves. (a) Calculations using the equivalent linearmethod, (b) using a fully nonlinear calculation approach.

developing Figure 6.10. The calculations plotted in Figure 6.11a were doneusing an equivalent linear method and those in 6.11b using a nonlinearapproach. The purpose of the calculations was to illustrate the effect of scaling

the input earthquake time history. Three motions are applied to the system: 0.3,1.0 and 3.0 times the original motion (that is every acceleration value ismultiplied by a scaling factor); thus there is a factor of 10 between the weakestand strongest motions. If the system behaved in a linear elastic manner theshapes of the response spectra would all be the same with the ordinates scaledaccording to the scale factors. Nonlinear behaviour will produce a differentresult, which both parts of Figure 6.11 demonstrate. First the peak responseoccurs at a period which is longer than the small strain period of the soil layer,0.27 second. Both parts of the figure show a progressive lengthening of theperiod at peak response as the scaling factor increases; at the scaling factor of0.3 the period of peak response is only just a little greater than 0.27 seconds.

Even more important, though, is the fact that the peak spectral accelerationdoes not increase in a linear manner with scaling. It is apparent, particularly forthe nonlinear results in Figure 6.11b that as the level of earthquake accelerationincreases the peak spectral acceleration does not increase in the sameproportion. Thus, for sufficiently severe earthquake motion, nonlinear soilbehaviour reduces the response at the surface of the soil. Figure 6.11, then, isdemonstrating the effect illustrated in Figure 6.7.

6.1.4 Two and three dimensional site response

The calculations discussed in section 6.1.3 are possible with modest computer

resources because the direction of wave propagation is restricted to vertical.However, actual earthquake motion will be more complex involving


14/34


152

propagation in two or three dimensions. This means that only when there is anabrupt change in layer properties can the one-dimensional model be useful.

Given the sophisticated computer resources presently available it is possible to

calculate routinely the elastic response of valleys and basins. In fact, thechallenge is about defining the three dimensional geometry of the soil depositand the wave velocities of the oils. Such studies provide useful insight, see forexample Bard and Bouchon (1985) and Rial et al (1992), which show thatnatural frequencies for two and three dimensional configurations are differentfrom the one-dimensional values. Thus it should not be expected thatearthquake response spectra will have peaks corresponding to the peaks inFigure 6.9a and 6.10c.

The nonlinear response of two and three dimensional systems can, in principle,also be calculated but such endeavours are well beyond engineering design

work.

6.2 EARTHQUAKE LOADING AND DESIGN RESPONSE SPECTRA

The ground motion measured during an earthquake can be expressed as amulti-directional random, or pseudo-random, signal of acceleration, velocityand displacement. The motion typically lasts for some tens of seconds to aminute or so. Design of buildings and foundations to resist these motions, inthe first instance, adopts a pseudo-static viewpoint in which a base shear andmoment are obtained from a loadings standard.

In NZ this information is obtained from the document: NZS 1170.5:2004"Earthquake Actions New Zealand. In Europe it is given in Eurocode 8:Design of Structures for Earthquake Resistance (2003). In the United Statesthe Uniform Building Code (1997) and more recently Federal EmergencyManagement Agency documents: FEMA 356 Prestandard and commentaryfor the seismic rehabilitation of buildings (2000) and FEMA 368 NEHRPRecommended provisions for seismic regulations for new buildings and otherstructures (2000), as well as other documents, deal with these actions. Asummary of the various Japanese requirements is given by Hamada et al (2000).

Despite this pseudo-static approach the frequency content of an earthquakemotion can be important. A typical earthquake record can be regarded as thesuperposition of motions with periods in the range of about 0.10 to 10.0seconds. The standard technique for obtaining information about thefrequency content of an earthquake is to use the so-called response spectrum.

This a plot showing the peak response generated by the ground motion. Theground acceleration time history is used to compute the response of a series ofsingle degree of freedom "structures" of differing periods. The calculations aredone for each period of the single degree of freedom model by computing theresponse for the entire earthquake time history; the peak response for thatperiod is then plotted as one point on the response spectrum. Clearly the

amount of computation required is considerable and requires a computer for


15/34


153

Figure 6.12 Concept of the Acceleration Response Spectrum (after Seed

and Idriss (1969).


16/34


154

evaluation When the period of one of these oscillators corresponds with aperiod that contains a significant portion of the earthquake energy then there

will be a peak in the response spectrum. In this way the response spectrum canbe considered as a representation of the frequency content of an earthquake.

The concept is illustrated in Figure 6.12.

The peak ground acceleration is read off from the response spectrum as thezero period ordinate. For the El Centro record, used as the basis of theillustration of the concept in Figure 6.12, this is a little in excess of 0.3g. Thepeak ground acceleration depends on the magnitude of the earthquake (amountofenergy released), distance of the site at which the recording is made from thehypocentre and, as discussed above, the nature of the ground conditions at thesite.

A loadings standard uses a number of terms in arriving at a design base shearfor a particular structure. The shear force at foundation level (base shear) isgiven by:

H = C(T)Wt (6.1)

where: H is the base shear,Wt is the vertical load (the sum of the permanent load plus the

appropriate imposed load).C(T) is a dimensionless spectral shape factor coefficient which is a

function of the structural period, T.

We will follow, for illustrative purposes, the provisions of NZS 1170.5 (2004)herein, although other loading standards follow similar approaches. Thecoefficient C(T) is a coefficient given by:

1 1 1( ) ( ) ( , )hC T C T Z RN T D (6.2)

where: T1 is the first mode period of the structureCh(T1) is the basic seismic hazard acceleration coefficient given in NZS

1170.5Z is the hazard factor (maps the expected distribution of seismic

severity across the country),

R is the return period factor,and N(T1,D) is the near fault factor (D is the distance to the fault).

An alternative approach, used particularly for non-standard structures, is tocompute the time history of the response of a numerical model of the structureto a design ground motion. From this the complete time history of thefoundation actions is obtained, in which case slightly different spectra are used.In this case the stiffness and capacity of the foundation elements needs to bemodeled along with the numerical model for the above ground part of thestructure. The challenge here is to develop an effective model of the wholestructure/foundation system.


17/34


155

6.2.1 New Zealand Design code seismic hazard acceleration coefficients

Categorization of recorded earthquake motions in the form of Figure 6.5 is thebasis of the seismic hazard acceleration coefficients in loading standards.

The form of the spectral shape factor curves in NZS1170.5 is given in Figure6.13 for elastic response. The four curves are for rock, stiff soil, deep soil and

very soft soil. These curves are used when the equivalent static method is beingfollowed in design. When the design is based on time history calculations theform of the curves is different at low periods. Additionally designers oftenintend a ductile response for the structure under the design earthquake. Thisinvolves a reduction in the spectral acceleration because the elastic spectrum isdivided by the ductility and a performance factor, which is less than unity, isalso applied. Figure 6.14 compares the elastic spectrum for a soil profile withthat for a ductility of three and a performance factor of 0.67. It is clear, that at

the expense of structural damage, designing the structure for ductility reducesthe earthquake actions in the structure and also on the foundation.

In the last 20 years or so earthquake events have been recorded at manyaccelerographs in the vicinity of the event. This has lead to betterunderstanding of the spread of earthquake motions and the attenuation of themotion with distance from the fault generating the motion. What has emergedfrom examination of these records is that the intensity of the motion varies notonly with distance from the causative fault but also with angular orientation

with respect to the fault plane, in particular the directions parallel to the faultand those normal to it. The distribution of energy is more concentrated along

the fault parallel direction that in the normal direction. Furthermore theshaking intensity is more severe for rupture towards a site than away from it,Somerville et al (1997), Abrahamson (2000) and Bray (2006). The near faultterm in equation 6.2 makes some allowance for these effects in specifying thedesign acceleration.


18/34


156

0 1 2 3 4 50

1

2

3

Period (sec)

Spectalshape

factor

Figure 6.13 Spectral shape factor curves for equivalent static design from

NZS1170.5 (2004). The curves are for rock, shallow soil, deep soil and very

soft soil.

0 1 2 3 4 50

1

2

3

Period (sec)

Spectralshapefactor

Figure 6.14 Comparison between the NZS 1170.5 elastic shallow soil curveand that for a ductility of 3 with a performance factor of 0.67.


19/34


157

0 1 2 3 4 50

0.1

0.2

0.3

Period (sec)

Spectraldisplacementshapefactor

Figure 6.15 NZS1170.5 (2004) spectral shape factor curves for displacement

based design obtained from the acceleration coefficient curves in Figure

6.13. The curves are for rock, shallow soil, deep soil and very soft soil.

6.2.2 EC8 seismic hazard acceleration coefficients

The way in which the seismic hazard curves work in Europe is explained byFardis et al (2005). EuroCode 8 has five ground classes, detailed herein in

Table 6.1, and Figure 6.16 gives the coefficient curves for these groundclasses. For these there are different seismic hazard curves as shown in Figure6.16. The curves presented in Figure 6.16 have the same general form as thosein NZS1170.5. However, EC8 gives two levels of these curves, those plotted inFigure 6.16 are for a so-called Type 1 EQ (magnitude > 6.5), there are separatecurves for Type II events (lesser magnitude events).


20/34


158

Table 6.1 Ground types defined in EC8


21/34


159

0 1 2 3 40

1

2

3

Period(seconds)

HazardCoefficient

Figure 6.16 Curves of seismic hazard coefficients specified in EC8 (Type 1

EQ, Ms > 5.5)

6.3 SOIL-STRUCTURE INTERACTION

In Figure 6.12 the single degree of freedom structures are shown as beingrigidly fixed to the ground. When a structure is founded on a soil profile thefoundation flexibility contributes to the response of the system. In using theseismic hazard coefficient curves, such as those in Figures 6.10 and 6.13, theperiod used is not simply that of the structure but includes the effect of thefoundation flexibility. This is know as soil-structure interaction and Figure 2.9illustrates how, because of soil-structure interaction, the foundation response isa little different from the free field response. Figure 6.8 illustrates how thebedrock motion is modified by the soil layers (site effect); this will be furthermodified by the interaction between the structure-foundation system with thesoil adjacent to the foundation.

6.4 WIND AND WAVE LOADING

For comparative purposes the following is some brief information about windand wave loading.

6.4.1 Wind loading

Peak wind loading conditions, like a major earthquake, are relatively rareoccurrence. As with earthquake loading the return period is also used when

specifying a design wind loading event.


22/34


160

Figure 6.17Wind loading spectrum for a tall building. (a) Translational and

(b) rotational modes. (after Dalgliesh (1982)).

Unlike earthquake loading, a structure will be subjected to wind loading, of less

than the maximum design levels, over the whole of the design life. Thus thefoundation must perform satisfactorily under very large numbers of lowintensity cycles of loading.

The frequency of wind loading becomes important for tall flexible structureswhich can be excited by the shedding of vortices. The flexibility of thefoundation system may contribute to or inhibit this effect. An example of the

wind loading spectrum applied to a very tall building is given in Figure 6.17,Dalgliesh (1982).

Comparison of Figures 6.4 and 6.9 with 6.14 illustrates the difference in

frequency content between wind and earthquake loading.

6.4.2 Wave Loading

Wave loading, like wind loading, occurs at a low level right through the designlife of an offshore structure. Severe events, which determine the maximumloadings, are rare just the same as is the design earthquake and the design windloading.

Wave loading is important in the design of harbour facilities and for wallsprotecting the shoreline against waves. However, the most significant offshore

structures are associated with the oil and gas industry. These are of two types:gravity platforms which sit on the sea bed and pile supported platforms. As


23/34


161

structures these are amongst the biggest that the civil engineer encounters,Figure 6.18 compares one of the North Sea gravity platforms with the New

York skyline.

Wind and wave loading also differ from an earthquake in the duration of anevent. Whereas a large earthquake is over in minute or so in duration a windor wave storm may continue for hours and even, in a very extreme event, fordays. The most significant difference between earthquake and wave loading is,once again, the frequency content. Wave periods, like the wind loadingperiods, are several seconds and longer. Figure 6.19 gives the results of wavemeasurements

Figure 6.18 Stratford B gravity structure compared with the United Nations

Building and the Manhattan skyline (after DiBiagio and Hoeg (1983)).

Figure 6.19Frequency content of waves in the North Sea (after Hassleman et

al (1973)).


24/34


162

in the North Sea off the northern coast of Germany reported by Hasselmannet al (1973). It is apparent that the wave activity peaks for periods in the 4 to5second range. The ordinate in the diagram is related to the wave heightFurther offshore in the open ocean wave heights in a severe storm will increase

and the periods lengthen from those shown in Figure 6.19.

6.5 LOAD AND RESISTANCE FACTOR DESIGN (LRFD)

In ultimate limit state design we require that the capacity of the system isgreater than the demand. By capacity we mean the strength that can bemobilised if all the shear strength of the soil is used. By demand we mean theactions that are applied to the foundation by the loads imposed by thestructure. However, things are not quite as simple as they sound here. Thecapacity is not an easily specified exact number because the soil properties are

variable hardly surprising for a natural material. Similarly exact values for the

loads are not easily specified. Thus our ultimate limit state design must takeaccount of two sources of uncertainty that of the ground properties and thatof the applied loadings. There are various approaches to ultimate limit statedesign; this section deals with the one in use in NZ.

In New Zealand and elsewhere (for example Barker et al (1991) and in Canada(Canadian Foundation Design Manual (2007)), Australia and North America)the Load and Resistance Factor Design (LRFD) approach is used. The basicequation is:

RiFi 6.3

where:Fi are the applied loads.R is the ideal1 resistance (or strength).

i are load factors.

is the strength reduction factor2 (Note that the usual symbol for

the strength reduction factor in limit state design is . As this isthe standard symbol for the friction angle in geotechnical

engineering the symbol is used herein.)

For the LRFD method the strength of the foundation is evaluatedwith the actual values for the shear strength parameters and then thestrength value reduced by the application of the strength reductionfactor (a number less that unity).

To some extent the decision as to which method of ultimate limitstate design to use depends on what is being followed in other areassuch as structural design. Structural design in New Zealand is basedon an LRFD approach. The load factors are given in the StructuralDesign Actions Standard (AS/NZS 1170:2002) and the various

1This resistance is commonly referred to in several ways. Among them are theoreticalresistance, nominal resistance as well as ideal resistance.

2Referred to as a Performance Factor or Resistance Factor in some documents.


25/34


163

materials standards give values for the strength reduction factors.This being the case it is attractive to follow a LRFD approach infoundation design so that all aspects of civil engineering design inNZ are done on the same conceptual basis.

6.5.1 Provisions for the ultimate limit state design of

foundations

A factored load and resistance approach is in two parts; (i) the loadfactors and load combinations, and (ii) the strength reductionfactors applied to the ideal strength and resistance.

AS/NZS 1170 uses the notation G for permanent actions, Q forimposed actions, and E for earthquake actions.

When calculating design actions in accordance with AS/NZS 1170the permanent load factor on active and Ko earth pressures is 1.5(clause 4.2.3f).

In many stability calculations the permanent load contributes to theresistance which is generated by mobilised soil strength. In this casea load factor on the permanent load of 0.9 is applicable inaccordance with clause 4.2.1 (a) of AS/NZS1170. This isinterpreted herein as applying to loads that come to the foundationfrom external sources. However, the unit weight of the soil,although a permanent load, is not factored. The factoring of the

unit weight of the soil is not followed in Eurocode 7 (EC7) as thecommittee drafting that document have marshalledseveralarguments suggesting that this factor should always be unity,Orr (1993c).

Table 6.2 Overview of factoring of actions for Ultimate Limit State

Geotechnical Design (AS/NZS 1170)

Load condition Load factorPermanent load only (G) 1.35 (AS/NZS

1170)

Permanent loads from soil weight 1.0Permanent stabilising loads other thansoil weight

0.9 (AS/NZS1170)

Loads from water pressures 1.2 (AS/NZS 1170)Imposed loads (Q) 1.5 (AS/NZS 1170)Loads from earth pressures 1.5Permanent load + Imposed load 1.2G + 1.5Q

(AS/NZS 1170)

Permanent load + Imposed load + EQload

1.0 (AS/NZS 1170.5)

Thus the application of the loads factors in geotechnical design is asfollows:


26/34


164

For bearing capacity calculations the unit weight of the soil is not factored by

0.9 in either the Nq or Nterms of the bearing capacity equation.For the design of foundations under static conditions the following

load factoring is recommended:

Active and Ko thrusts are factored by 1.5 when determiningthe required proportions for the wall. In calculating thesethrusts the unit weight of the backfill is not also treated as apermanent load and factored. (Note that passive thrusts arepart of the resistance and so are not treated with loadfactors.)

The weight of the wall acts to promote the stability bygenerating sliding resistance. However, many retaining wall

applications the weight of the wall behaves in the samemanner as the weight of the backfill, for example a cantilever

wall, because of this the weight of the wall is not factored by0.9.

Any additional permanent loads which assist stability, egfrom an adjacent building which will be in place for thedesign life of the foundation, are factored by 0.9.

If there is some surcharge in front of the wall, even thoughthis is considered as contributing to the resistance, the unit

weight of this is not factored by 0.9.

For the design of pile foundations dragdown loads arefactored as a static permanent load.

In AS/NZS 1170 imposed load reduction factors arespecified (clause A3.1.1 in the standard). These allow for thefact that not all parts of a structure will have full imposedloading acting simultaneously. For example the design of afloor slab for an office would be designed for full imposed

load as would the beams supporting that part of the floorslab. But for a multi-storey building the foundation designwould not assume that full imposed load was carried byevery square metre of each floor and that these are summedover the number of floors to get the foundation actions. Soin estimating foundation design actions one applies all thepermanent load from the structure and a portion of thedistributed imposed load.


27/34


165

6.5.2 Earthquake actions for ultimate and serviceability limit

states

In this course we will only touch on earthquake actions on

foundations, but for completeness the following comment isincluded. Earthquake actions are specified in NZS 1170.5 under twoheadings: Eu for ultimate limit state and Es for serviceability limitstates. The ultimate limit state actions are evaluated as set-out insection 3.1 and 5.2.1 of NZS 1170.5. These estimates may includethe possibility that the structure will respond to the earthquake in aductile manner. For the serviceability limit state the structure isassumed to respond elastically and the actions are evaluated for thelargest translational period of vibration of the structure, followingagain what is specified in section 3.1 of NZS 1170.5. But in additiona value of 0.7 for the Structural Performance Factor, clause 5.2.1.2

of NZS 1170.5, is applied to the actions.

Evaluating the response of a foundation to serviceability limit stateearthquake actions requires estimating the immediate cyclicdeformations induced by the earthquake actions on the foundations(we will see in Chapter 10 how this can be done by assuming elasticbehaviour of the soil beneath the foundation, undrained forcohesive soils). These cyclic deformations are then superimposed onthe static deformations of the foundation (usually settlement).

Combinations of ac tions for ultimate limit states

Furthermore the factors applied to actions are adjusted when thereare combinations of actions applied, for example: permanentactions, imposed actions and earthquake actions, or permanentactions, imposed actions and wind actions.

Table 4.1 of AS/NZS 1170.0 gives the combination factor,

c, as 0.4.

Section 4.2 of NZS 1170.5 gives the combination factor

when earthquake loads are involved, c_EQ, as 0.3.

Some examples of combinations, fleshing out those in clause 4.2.2of NZS 1170.5, are:

ULS Permanent action only: 135. G

(Note that for this case we have an area reduction factor (ifapplicable) and long term reduction factor (if applicable), but thereis no combination factor.)

ULS Permanent plus short term imposed action:12 15 a. G . Q

ULS Permanent plus long term imposed action:

12 15 a. G . Q


28/34


166

ULS Permanent, imposed and wind actions: c a uG Q W

ULS Permanent, imposed and earthquake actions:

c _ EQ a uG Q E

SLS permanent and long term imposed actions:

a lG Q

SLS Permanent, imposed and earthquake actions: a c _ EQ sG Q E

6.5.3 Strength Reduction Factors for Shallow Foundations

In New Zealand the Structural Design Actions Standard specifies

how actions are to be estimated and also how they are factored onthe right hand side of inequality 6.3. The strength reduction factorsfor the left hand side of inequality 6.3 are specified in the variousmaterials standards.

Table 4.5 gives values of for the design of shallow foundations

and retaining walls (bc is for the bearing capacity and sl for thesliding ultimate limit state).

Sliding resistance depends directly on the tangent of the frictionangle. For passive pressures and bearing capacity the resistance isnot controlled directly by the tangent of the friction angle but bycomplex functions which change very rapidly once the friction angleis beyond the mid 30's. In addition the displacements required tomobilise passive resistance and bearing capacity are large.

Table 6.3 Shallow foundation strength reduction factors

Strength reduction factors for the ultimate limit state in bearing (bc)and for ultimate limit states involving passive earth pressure ( )

Load combination Strength reduction factor rangeLoad combinations including

earthquake overstrength

0.80 - 0.90

All other load combinations 0.45 - 0.60

Strength reduction factors for sliding of shallow foundations (sl)Load combination Strength reduction factor rangeLoad combinations includingearthquake overstrength

0.80 - 0.90

All other load combinations 0.80 - 0.90

Tentative strength reduction factor for embedded retaining walls(ew)All load combinations 0.80


29/34


167

Table 6.4 Strength reduction factors for deep foundation design

Range of Values of cStatic load testing to failure 0.65 - 0.85

Static proof (not to failure) load testing 0.70 - 0.90Static analysis using CPT data 0.45 - 0.65Static analysis using SPT data in cohesionless soils 0.40 - 0.50Static analysis using laboratory data for cohesivesoils

0.45 - 0.55

Measurement during installation of proprietarydisplacement piles, using well established in-houseformulae

0.50 - 0.65

Load combinations including earthquakeoverstrength

0.80 - 0.90

In this chapter the primary applications intended for the ultimatelimit state method are retaining walls and shallow foundations.

However, values of values similar to those for shallowfoundations are given by Barker et al (1991) for deep foundations.In addition Becker (1993) gives strength reduction factors for deepfoundations. The draft of AS 2159 (1993) gives a range of values for

pc ( pile capacity), Table 4.6 is modelled along the same lines.

6.5.4 Strength reduction factors for aseismic design

Taylor (1976) and Taylor and Williams (1979) explain, within thecontext of the factor of safety approach, that design bearingcapacity factors of safety need to be reduced to account for the useof factored loads. The value to be used depends on the particularload combinations under consideration. Notwithstanding thecomplication introduced by the use of various load combinations

Taylor suggested that design should proceed on the basis that theoverall factor of safety (including load factoring) for staticconditions should be about 3, about 2 for non-capacity designearthquake loading, and 1.1 for capacity designed foundations.

For earthquake loading the values given in Table 4.5 are used asthe earthquake load combinations in AS/NZS 1170 have unit load

factors (so the values are roughly equivalent to a factor of safety

of 2). For capacity loading a value of 0.9 is equivalent to theprevious recommendation of a factor of safety of 1.1.

Aseismic design of foundations is considered under two distinctheadings. There are foundations for capacity designed structuresand those for structures which are not designed for capacity actions.For capacity design special actions are evaluated based on theoverstrength capacity of the structure, these actions give themaximum loads that the structure could ever apply to thefoundation. When capacity design principles are not followed the


30/34


168

ultimate capacity load combination consists of permanent loads plusimposed load plus earthquake loads, that is the load factors areunity.

6.6 PARTIAL FACTOR OF SAFETY APPROACHES

The Partial Factor of Safety approach is used in Europe, Eurocode7 (EC7, 2001), Orr (1993), Simpson and Driscoll (1998).

The traditional total factor of safety approach lumps the twosources of uncertainty into one parameter, the factor of safety. Thepartial factor of safety approach handles the uncertainly in theloading by applying partial factors to the applied actions (generallygreater than unity) and the uncertainty in the shear strength of the

soil by applying partial factors to the shear strength properties. Inthe case of the partial factor method the foundation strength isevaluated using reduced values for the cohesion and friction angle.

iFi 6.4

where: is the resistance calculated using the design strengthparameters obtained by reducing the characteristic

strength values with partial factors of safety (ie (c)d =

c/Pfsc and (tan)d = tan/Pfs, or (su)d = su/Pfsu).Pfsc etc are the partial factors of safety. (Note that there is

more than one partial factor of safety.)c etc are the characteristic strength values (defined in EC7:

The characteristic value is a cautious estimate of themean value.)

(c)d etc are the design values of the strength parameter foruse with the partial factor of safety method.

The Partial factor approach has some appeal from the geotechnical point of viewin that the variability of frictional and cohesive shear strength parameters isdifferent, Figures 4.48, so that different partial factors can be applied to these

strength parameters.

6.7 GENERAL COMMENTS ABOUT BUILDING FOUNDATION

CONCEPTUAL DESIGN

Eurocode 8, Fardis et al (2005), has a list of suggestions under the heading of conceptualdesign, considerations that are thought likely to produce more robust structures. Theitems listed are:

structural simplicity

uniformity, symmetry and redundancy


31/34


169

bi-directional resistance and stiffness

torsional resistance and stiffness

diaphragmatic behaviour at the storey level

adequate foundations

Although the foundation is covered in the last bullet point, all the points relate tofoundation design. For example the structure is expected to have bi-directionalresisstance and stiffness (ie the direction of earthquake attack is not known so thestructure needs to have resistance available in all directions). Clearly the sameconsiderations should be applied to foundation design.

The book by Arnold and Reitherman (1982) provides many examples of buildingconfigurations intended to provide robust performance during earthquake shaking.

They, too, make the point about regularity of configuration whenever possible.

It is regarded as good practice that foundation elements be tied together. The thinkingbeing that if for some reason one foundation element is in distress then the actions canbe passed along the tie beam to adjacent foundation elements. An example of thispractice, from Christchurch, is shown in Figure 6.20 from which it is seen that the tiebeams are quite substantial.

It is of note that part V of Eurocode 8 covers geotechnical matters and foundations.

Figure 6.20 Building founded on piles with tie-beams between which became

exposed following settlement of the upper parts of the soil profile during the February

22, 2011 earthquake.


32/34


170

REFERENCES

Abrahamson, N. A. (2000) State of the practice for seismic hazard evaluationProc. GeoEng2000, Melbourne. Invited paper for the Earthquake

Geotechnical Engineering Special Session, 27 pages.Aki, K. (1988) Local site effects on strong ground motion. Proc. ASCE

Specialty Conf. Earthquake Engineering and Soil Dynamics II, pp. 103-155.

Aki, K. and Irikura, K. (1991) Characterisation and mapping of earthquakeshaking for seismic zonation. Proc. 4th. International Conference onSeismic Zonation, Stanford. Vol. 1, pp. 61-110.

ANZ/NZS 1170: 2002 Structural design actions. StandardsAustralia/Standards New Zealand. Sydney, Wellington.

Arnold, C and Reitherman, R (1982) Building Configuration and Seismic Design,Wiley, New York.

Bard, P-Y. and Bouchon, M. (1985) "The two-dimensional resonnce of sedimentfilled valleys". Bull. Seismological Society of America. Vol. 75, No. 2, pp.519-541.

Barker, R. M., Duncan, J. M., Rojani, K. B., Ooi, P. S. K., Tan, C. K. and Kim, S.G. (1991) "Manuals for the design of bridge foundations", Report 343National Cooperative Highway Research Program, TransportationResearch Board, National Research Council, Washington DC.

Becker, D. E. (1996) Eighteenth Canadian Geotechnical Colloquium: Limitstated design for foundations. Part II. Development for the NationalBuilding Code of Canada. Canadian Geotechnical Journal, Vol. 33, pp.984-1007.

Borcherdt, R. and Donovan, N. (1989) "Loma Prieta earthquake October 17,1989: Geosciences", Preliminary Reconnaissance Report, EERI, pp. 5-22.

Bray, J D and Rodriguez, A (2006) Design ground motions near active faults,New Zealand Workshop on Geotechnical Earthquake Engineering 2006,University of Canterbury, pp. 19-31.

Building Industry Authority (1999) B1/VM4 Foundations. Building IndustryAuthority, Wellington.

Canadian Geotechnical Society (2006) Canadian Foundation EngineeringManual, 4th edition, BiTech, Richmond, British Columbia.

Chang, C.Y., Mok, C.M., Power, M.S., Tang, Y.K., Tang, H.T., and Stepp, J.C.(1990) Equivalent linear versus nonlinear ground response analyses at

the Lotung seismic experiment site. Proc. 4th. U.S. National Conf. onEQ Engineering, Palm Springs, Vol. 1, pp. 327-336.

Chang, C.Y., Power, M.S., Tang, Y.K. and Mok, C.M. (1989) Evidence ofnonlinear soil response during a moderate earthquake. Proc. 12th.International Conference on Soil Mechanics and FoundationEngineering, Rio de Janeiro, Vol. I, pp. 1927-1930.

Chin, B-Y, Aki, K. and Martin, G.R. (1991) The nonlinear response of soil sitesduring the 1989 Loma Prieta Earthquake, EOS, 72, 339

Chin, B-Y, and Aki, K. (1991) Simultaneous study of the source, path and siteeffects on strong ground motion during the 1989 Loma Prietaearthquake: a preliminary result on pervasive nonlinear site effects, Bull.

Seis. Soc. of America, Vol. 81, No. 5, pp. 1859-1884.


33/34


171

Dalgliesh, W. A. (1982) "Comparison of model and full scale tests of theCommerce Court building in Toronto", Proc. Int. Workshop on Wind

Tunnel Modelling for Civil Engineering Applications, Cambridge, pp.575-589.

Darragh, R.B. and Shakal, A.F. (1991) The site response of two rock and soilstation pairs to strong and weak ground motion. Bull. Seis. Soc. of

America, Vol. 81, No. 5, pp. 1885-1899.DiBiagio, E and Hoeg, K (1983) "Instrumentation and performance observations

offshore", Proc. ASCE Conference on Geotechnical Practice in OffshoreEngineering, Austin, pp. 604-625.

Dickenson, S.E., Seed, R.B., Lysmer, J.L. and Mok, C.M. (1991) Response ofsoft soils during the 1989 Loma Prieta earthquake and implications forseismic design criteria. Proc. Pacific Conf. on EQ Engineering,

Auckland, Vol. 3, pp. 191-203.EC7 Drafting Committee (2001) Eurocode 7, part 1: Geotechnical Design,

General Rules. Comit Europen de Normalisation, Final draft,October 2001.

Fardis, M N, Carvalho, E, Elnashai, A, Faccioloi, E, Pinto, P and Plumier, A(2005) Designers Guide to EN 1998-1 and EN 1998-5 Eurocode 8:Design of structures for earthquake resistance. General rules, seismicactions, design rules for buildings, foundations and retainingstructures. Thomas Telford, London.

FEMA 356 (2000b) Prestandard and commentary for the seismicrehabilitation of buildings, Federal Emergency Management Agency,

Washington DC.FEMA 368 (2000a) NEHRP Recommended provisions for seismic

regulations for new buildings and other structures, Federal EmergencyManagement Agency, Washington DC.Hamada, M., Unjo, S., Nishimura, A., Uwabe, T., Une, S. and Yamakawa, H.

(2000) Earthquake Resistant Design Codes in Japan, Japan Society ofCivil Engineers.

Hashash, Y.M.A, Groholski, D.R., Phillips, C. A., Park, D, Musgrove, M.(2011) DEEPSOIL 5.0, User Manual and Tutorial, 107 p.

Hasslemann, K. et al (1973) "Growth and swell decay during the joint North SeaWave Project", UDC 551.466.31; ANE German Bight, DeutschesHydrographisches Institut, Hamburg.

Hryciw, R.D., Rollins, K.M., Homolka, M., Shewbridge, S.E., AND McHood, M

(1991) Soil amplification at Treasure Island during the Loma PrietaEarthquake, Proc. Second International Conference on RecentAdvances in Geotechnical Earthquake Engineering and Soil Dynamics,St Louis, pp. 1679-1685.

Idriss, I M (1990) Response of soft soil sites during earthquakes, Proc SeedMemorial Symposium, Berkeley, pp 273-289.

Idriss, I. M. (1985) "Evaluating seismic risk in engineering practice", Proc. 11th.ICSMFE, San Francisco, Vol. 1, pp. 255-320.

Idriss, I.M. (1990) Response of soft soil sites during earthquakes, Proc. SeedMemorial Symposium, pp. 273-289.

Jarpe, S.P., Cramer, C.H., Tucker, B.E. and Shakal, A.F. (1988) A comparison

of observations of ground response to weak and strong ground motion


34/34


at Coalinga, California. Bulletin Seismological Society of America, Vol.78 No. 2, pp. 421-435.

Jarpe, S.P., Hutchings, L.J., Hauk, T.F. and Shakal, A.F. (1989) Selected strong-and weak-motion data from the Loma Prieta earthquake sequence,

Seismological Research Letters, Vol. 60 No. 4, pp. 167-176.Kramer, S. L. (1996). Geotechnical Earthquake Engineering. Prentice Hall, New

Jersey.Meyerhof, G. G. (1993) Development of Geotechnical Limit state design,

Keynote Address: International Symposium on Limit State Design inGeotechnical Engineering, Copenhagen. Danish Geotechnical Society,

Vol. 1, pp. 1-14.NZS 1170.5: 2004 Structural Design Actions Part 5: Earthquake Actions

New Zealand Standards New Zealand, Wellington.Orr, T. L. L (1993) "Partial safety factors in geotechnical design", Paper

presented to the Geotechnical Society of Ireland, 9th November 1993.

Rial, J. A., saltzman, N. G. and Ling, H. (1992) "Earthquake-induced resonancein sedimentary basins". American Scientist, Vol. 80, pp. 566-578.

Romo, M.P., Ovando-Shelley, E., Jaime, A. and Herandez, G. (1989) Local siteeffects on Mexico City ground motions., Proc. 12th. InternationalConference on Soil Mechanics and Foundation Engineering, Rio de

Janeiro, Vol. 3, pp. 2001-2008.SANZ (1984) "NZS 4203:1984 Code of practice for general structural design and

design loadings for buildings", Standards Association of NZ.Schnabel, P. B., Lysmer, J. and Seed, H. B. (1972) SHAKE, Report No. EERC

72-12, University of California, Berkeley.Seed, H. B. (1969) "The influence of local soil conditions on earthquake damage",

Proceedings of Specialty Session No. 2 Soil Dynamics, 7th. ICSMFE,Mexico City, pp. 33-66.Seed, H. B. and Idriss, I. M. (1982). Ground Motions and Soil Liquefaction during

Earthquakes, Earthquake Engineering Research Institute, Berkeley.Simpson, B. and Driscoll, R. (1998) Eurocode 7 a commentary. Construction

Research Communications, London.Somerville, P. G., Smith, N. F., Graves, R. W. and Abrahamson, N. A. (1997)

Modification of empirical strong ground motion attenuation relationsto include amplitude and duration effects of rupture directivity,Seismological Research Letters, Vol. 68, pp. 199-222.

Taylor, P. W. (1976) "Code provisions related to soils and foundations", Bull.

NZ. Nat. Soc. Earthquake Engineering, Vol. 9 No. 1, pp. 68-73.Taylor, P. W. and Williams, R L (1979) Foundations for capacity designedstructures, Bull. NZ Nat. Soc. Earthquake Engineering, Vol. 12 No. 2,pp. 101-113.

Towhata, I. (2008) Geotechnical Earthquake Engineering. Springer, Berlin.Whitman, R. V. (1986) "Are the soil depositions in Mexico City unique", Proc.

ASCE Conference: The Mexico City Earthquakes - 1985 Factors involvedand Lessons learned. Mexico City. pp. 163-177.

EQ Design Actions

Documents

Transcript of EQ Design Actions