Economy in Steel

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Know what influences the cost of steel buildings and bridges and get dozens of practical suggestions that will

help you cut cost in steel design and construction without sacrificing quality.

This feature is adapted from two major sources:

Cost-Effective Steel Building Design -- The U.S. Approach, a paper in the journal Progress in

Structural Engineering and Materials by Charles J. Carter, Thomas M. Murray and William A.

Thornton. Charles J. Carter, SE, PE, is Director of Engineering and Continuing Education, American

Institute of Steel Construction, Chicago, IL, USA. Thomas M. Murray, PE, PhD, is Montague-Betts

Professor of Steel Design, Virginia Polytechnic Institute and State University, Blacksburg, VA, USA.

William A. Thornton, PE, PhD, is President, Cives Engineering Company, Roswell, GA, USA.

A Guide to Economical Practices in Steel Design and Construction, a compendium of

recommendations that was published by the Structural Steel Fabricators of New England (SSFNE).

It is presented in 6 sections as follows:

The Economy Equation -- In this feature, the factors in the cost of steel building construction in the

United States are discussed, with emphasis on the dominant components of the total cost.

Ways to Save Time and Money -- More than 40 suggestions that you can use in your office practice

today to work smarter, not harder -- and to improve the economy of steel building construction.

More Ways to Save Time and Money -- More than 40 more suggestions.

Recent Developments -- What's on the horizon and what will be its impact?

Future Trends and Needs -- Where do we need to go?

References -- An excellent source of useful (and practical) supplemental reading material.

You can also find some similar and addition information at the following links in AISC's Modern Steel

Construction magazine:

Economy in Steel -- Practical information for designers. From the April 2000 issue.

New Design Developments -- What's on the Horizon for Steel. From the April 2000 issue.

Reducing Fabrication Costs -- Ideas from the Field: Every fabricator and erector has a long list of ideas

on what design engineers can do to reduce the cost of building a steel structure. What follows are some

thoughts gathered in the field. From the April 2000 issue.

Reducing Fabrication Costs in Steel Bridges -- New software under development. From the April 2000

issue.

Reducing Joist Costs -- Advice from the Steel Joist Institute. From the April 2000 issue.

Value Engineering For Steel Construction -- A complete design is the best assurance that those who

must use that design will accurately interpret the intent of the designer. From the April 2000 issue.

The Economy Equation

The famous bank robber Willie Sutton was once asked a simple question: why do you rob banks, Willie? His

simple response: "because that's where the money is." Sarcastic? Maybe, but his answer showed why he was

so successful. The simple question in your mind right now is probably "what does Willie Sutton have to do

Economy in Steel -- A Practical Guide http://www.engr.psu.edu/ae/steelstuff/economy.htm

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with steel economy?" Well, like he said,

If you want to save money in steel construction, go where the money is!

To find where the money is, let's take a look at how costs are determined.

When a steel fabricator prepares a cost estimate for a typical project, the following steps are common:

Perform a detailed material and labor takeoff.

Weigh and price all materials, including waste materials, for which payment is based upon weight, such

as structural shapes, plates and bolting products.

Add the cost of supplemental materials for which payment is not based upon weight, such as welding

and painting products.

Add the cost of fabrication labor required for each operation, including overhead and profit.

Add the cost of all outside services required, such as pre-fabrication materials preparation, galvanizing,

shipping and erection.

Add the cost of shop drawings.

Add the cost of buyout items, such as steel deck and steel joists.

Evaluate the risk and need for contingency and add the appropriate amount.

Factor in schedule requirements and add the appropriate amount.

All of the components of the total cost identified in the foregoing estimating process can be classified into one

of four categories:

Material costs: This category includes the structural shapes, plates, steel joists, steel deck, bolting products,

welding products, painting products, and any other products that must be purchased and incorporated into the

work. It also includes the waste materials, such as short lengths of beams (called "drops") that result when

beams are cut to the specified length. By an order of magnitude, the most influential component of these

products on the total material cost of a building structure is the weight of the structural shapes.

As illustrated in Figure 1, the typical material cost has dropped in recent years from 40 percent of the total

cost in 1983 to 26 percent in 1998. This represents a 35 percent decline in material cost over the last 15 years.

Fabrication labor costs: This category includes the fabrication labor required to prepare and assemble the

shop assemblies of structural shapes, plates, bolts, welds and other materials and products for shipment and

subsequent erection in the field. It also includes the labor associated with shop painting. The total fabrication

labor cost is simply the cost of the shop time required to prepare and assemble these components, including

overhead and profit.

As illustrated in Figure 1, the typical fabrication labor cost has increased slightly in recent years from 30

percent of the total cost in 1983 to 33 percent in 1998. This represents a 10 percent increase in fabrication

labor costs over the last 15 years.

Erection labor costs: This category includes the erection labor required to unload, lift, place and connect the

components of the structural steel frame. The total erection labor cost is simply the cost of the field time

required to assemble the structure, including overhead and profit.

As illustrated in Figure 1, the typical erection labor cost has increased in recent years from 19 percent of the

total cost in 1983 to 27 percent in 1998. This represents a 42 percent increase in erection labor costs over the

last 15 years.

Other costs: This catch-all category includes all cost items not specifically included in the three foregoing

categories, including outside services other than erection, shop drawings and the additional costs associated

with risk, the need for contingency, and the schedule requirements of the project.

As illustrated in Figure 1, the typical cost in this category has increased slightly in recent years from 11

percent of the total cost in 1983 to 13 percent in 1998. This represents an 18 percent increase in other costs


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over the last 15 years.

Figure 1. Material, shop labor, erection labor and other costs; 1983 through 1998.

Obviously, very few projects, designers, fabricators and erectors are exactly alike. Given this, the exact

distribution of the total cost among these four categories can and will vary based upon the specific

characteristics of a given project, including the design and construction team. In some specialized cases, any

one of the four cost centers may dominate the total cost. Nonetheless, it can be stated that the current

distribution of cost, rounded to the nearest 5-percent increment, among these four centers for a typical

structural steel building is approximately as follows:

Material costs -- 25%

Fabrication labor costs -- 35%

Erection labor costs -- 25%

Other costs -- 15%

Cost Conclusion:

Thus, in today’s market, labor in the form of fabrication and erection operations typically accounts for

approximately 60 percent of the total constructed cost. In contrast, material costs only account for

approximately 25 percent of the total constructed cost. Clearly then, least weight does not mean least cost.

Instead, project economy is maximized when the design is configured to simplify the labor associated with

fabrication and erection. Willie Sutton would go after the labor.

Ways to Save Time and Money

Given The Economy Equation above, the following are 40 basic suggestions that you can use in your office

practice today to work smarter, not harder -- and to improve the economy of steel building construction.

Communicate!

With the division of responsibilities for design, fabrication and erection that is normal in current U.S. practice,

open communication between the engineer, fabricator, erector and other parties in the project is the key to

achieving economy. In this way, the expertise of each party in the process can be employed at a time when it

is still possible to implement economical ideas. The sharing of ideas and expertise is the key to a successful

project.

Take advantage of a pre-bid conference.

When in doubt about a framing detail or construction practice, consult a knowledgeable fabricator and/or

erector. Most will gladly make themselves available at any stage of the game for a pre-bid conference, such as

to help with preliminary planning or discuss acceptable and economical fabrication and erection practices. A

pre-bid conference can also be used to communicate the requirements and intent of the project to avoid


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misunderstandings that can be costly. Many times, fabricators and erectors can provide valid cost-saving

suggestions that, if entertained, can reduce cost without sacrificing quality.

Issue complete contract documents, when possible.

Design drawings and specifications are the means by which the owner, architect and/or engineer

communicates the requirements for structural steel framing to the fabricator and erector. Care in preparing

these and other contract documents is important, not only to assure responsive bids or estimates, but also to

minimize the potential for misrepresentation, errors and omissions in both bidding and the final product. The

most clear, complete and accurate design drawings and specifications will generally net the most accurate and

competitive bids. Certainly they are the starting point for economical, timely construction in steel. For

guidance on what constitutes complete contract documents, consult the AISC Code of Standard Practice,

particularly Section 3 therein. When the nature of the project is such that it is not possible to issue complete

contract documents at the time of bidding, clearly provide the scope and nature of the work as far as what the

framing will be and what kinds of connections are required.

Don't forget to include the basics.

Show a North arrow on each plan. Show a column schedule. Include "General Notes" that cover the

requirements for painting, connections, fasteners, etc. in a manner that is consistent, complementary and

supplementary to the specification.

Late details can cost a lot.

Even simple detail items like roof- or floor-opening frames can cost a small fortune if delayed, particularly

when the delay forces installation after the steel deck is in place.

Show all the structural steel on the structural design drawings.

As indicated in the AISC Code of Standard Practice, structural steel items should be shown and sized on the

structural design drawings. The architectural, electrical and mechanical drawings can be used as a supplement

to the structural design drawings, such as by direct reference to illustrate the detailed configuration of the steel

framing, but the quantities and sizes should be clearly indicated on the structural design drawings.

Make sure the general contractor or construction manager clearly defines responsibilities for

non-structural and miscellaneous steel items.

Structural and non-structural steel items are identified in AISC Code of Standard Practice Section 2. Many

items, such as loose lintels, masonry anchors, elevator framing, and precast panel supports, could be provided

by more than one subcontractor. Avoid the inclusion of such items in two bids by clearly defining who is to

provide them.

Avoid "catch-all" specification language.

Language like "fabricate and erect all steel shown or implied that is necessary to complete the steel

framework" probably sounds good to a lawyer, but it really does not add much to quality or economy because

it is nebulous and ambiguous. What is implied? Such language probably results only in arguments, contingency

dollars or change orders -- and legal fees.

Avoid language that is subject to interpretation.

Vague notations, such as "provide lintels as required", "in a workmanlike manner", "standard" and "to the

satisfaction of the engineer" are subject to widely varying interpretations. Instead, when required, specify

measurable performance criteria that must be met.

Use standard tolerances.

ASTM A6/A6M defines standard mill practice. The AISC Code of Standard Practice defines fabrication and

erection tolerances. The RCSC Specification covers bolting acceptance criteria. AWS D1.1 establishes weld

acceptance criteria. These and other documents provide standard tolerances that are acceptable for the


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majority of cases. Generally, they present the most efficient practices.

In some cases, more restrictive tolerances may be contemplated for compatibility with the systems and

materials that are supported by the structural steel frame. Or tolerances may need to be defined for highly

specialized systems or when steel and concrete systems are mated. All non-standard practices should be cost

justified.

Specify paint only when it is needed.

Corrosion resistance for architectural or structural purposes may be an important criterion in the performance

of structural steel. Often, however, the actual conditions of use do not warrant extensive surface preparation

or shop painting, and in these cases no special surface preparation or treatment should be called for. For

example, steel that is enclosed in building finishes, fireproofed, or to be in contact with concrete generally

need not be painted. Furthermore, if a finish coat is not specified, a shop primer coat need not be specified as

it is of only minor influence on corrosion in the construction phase. For more information, see AISC

Specification Section M3 and it's Commentary.

When painting is necessary, don't ask too much of the shop coat.

The standard shop coat is usually about one mil thick, but can commonly vary up to two mils thick.

Recognizing that it is very difficult to apply more than two mils of dry paint thickness in one coat without

runs, sags or drips, it should also be recognized that shop-coat thicknesses in excess of two mils will generally

require multiple coats and substantially increase shop painting costs. Ultimately, it must be recognized that the

shop coat of paint (primer) is temporary and will provide minimal protection during exposure for steel that is

to receive a finish coat in the field. Depending upon the duration and nature of the exposure, the shop coat

may last from several weeks to many months. Regardless, the durability of the shop coat should be considered

if an extended delay in the application of the finish coat is anticipated.

Also, when painting is necessary, select the right surface preparation and paint system for the job.

The three most commonly used surface preparations are SSPC SP-2 (hand-tool cleaning), SSPC SP-3

(power-tool cleaning) and SSPC SP-6 (commercial blast cleaning). SSPC SP-2 or SP-3 cleaning is usually

satisfactory for an ordinary shop prime coat. If conditions call for a high-performance paint system for long

term, low maintenance protection, SSPC SP-6 is more frequently required. When assemblies are to be blast

cleaned, consider the limitations on size and length, which vary depending upon the available equipment.

Careful consideration should also be given to specifying a paint system that will satisfy the required degree

of corrosion protection. A high-quality paint system (or galvanizing) can be cost-effective or even essential for

certain applications, as in open parking structures. In these cases, life-cycle costing should be performed. An

alternative to high-quality surface treatments, in normal atmospheric environments, may be ASTM A588

(weathering) steel.

Watch out for primer/fireproofing incompatibility.

For steel that is to receive spray-applied fire-protection, the fire protection manufacturer or applicator should

be consulted to determine their recommendations and/or preferences for painted or unpainted steel. If a

painted surface is preferred, the paint should be compatible with the fire protection.

Above all, make sure to coordinate with the architect so that the primer and finish coat are compatible!

The designer should consult with the paint manufacturers to ensure that shop-coat primers and finish-coat

paints are compatible. Incompatible coatings are all too common.

When specifying galvanized members, keep the maximum lengths in mind.

Galvanizing dip tanks are generally limited to a member length of 40 ft. Longer members often can be double-

dipped, as long as the noticeable zone of overlap between the two dips into the tank is not objectionable.

Clearly state any inspection requirements in the contract documents.

The scope and type of inspection of structural steel should be indicated in the project specification. Make sure


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that the requirements for inspection are appropriate for the application. For example, the inspection of groove

welds that will always be in compression during their service life is probably not required. Also, make sure

shop inspection is scheduled so that it does not disrupt the normal fabrication process.

Avoid the use of brand names when specifying common products.

When many manufacturers make a product, or there are acceptably equivalent products, avoid specifying the

product by brand name. When it is necessary to indicate a brand name for the purposes of description, be sure

it is a current, readily available product. Whenever possible, allow the substitution of an "equal". One

excellent example: paint.

Try to avoid them entirely, but when you can't, clearly identify changes and revisions.

Changes and revisions that are issued after the date of the contract generally have some cost associated with

them. For example, material may have already been ordered, shop drawings may have already been drawn

and shipping pieces may have already been fabricated. Thus, it is best to avoid a default reliance on the

change and revision process as a means to expedite schedules. However, when changes or revisions are

necessary or desirable, they should be clearly identified so that all parties can recognize them and account for

them.

Provide meaningful and responsive answers to requests for information.

When the fabricator asks for a design clarification through an RFI, the most prompt and complete response,

within the limitations of the available information, will be beneficial to all parties. If the RFI involves

information on a shop drawing approval submission, it is best to provide the most specific answer possible. Try

to avoid responses such as "architect to supply", "general contractor to supply", or "verify in field".

Use 50 ksi steel in wide-flange member design.

U.S. wide-flange steel shape production today is normally 50 ksi by default. Specifying ASTM A992, ASTM

A572 grade 50 or ASTM A529 grade 50 for W-shapes is actually now less expensive than specifying ASTM

A36 material, as explained here. Thus, even if deflection, drift, vibration or minimum-size criteria control, the

higher-strength material will still often help at little or no added cost for bolt bearing and block shear rupture

strength as well as in the elimination of detail material like transverse stiffeners and web doubler plates. For

further information, see Carter.

Use 36 ksi steel for plates and angles.

ASTM A36 material is still predominant in angles; so much so that it is difficult to obtain 50 ksi angle material,

except by special order from the rolling mill. Additionally, there is still a cost differential between 36 ksi and

50 ksi plate products.

Consider the use of hollow structural sections (HSS).

Square and rectangular HSS are available in ASTM A500 grades B and C with 46 and 50 ksi yield strengths,

respectively. Round HSS are available in ASTM A500 grades B and C with 42 and 46 ksi yield strengths,

respectively. Although their material cost is generally higher, HSS generally have less surface area to paint or

fireproof (if required), excellent weak-axis flexural and compressive strength, and excellent torsional

resistance when compared with wide-flange cross-sections.

Be careful when specifying beam camber.

Don’t specify camber below ¾-in.; small camber ordinates are impractical and a little added steel weight may

be more economical anyway. Also, do not overspecify camber. Deflection calculations are approximate and

the actual end restraint provided by simple shear connections tends to lessen the camber requirement.

Consider specifying from two-thirds to three-quarters of the calculated camber requirement for beams

spanning from 20 to 40 ft, respectively, to account for connection and system restraint. In any case, watch out

when rounding up the calculated camber ordinate, particularly with composite designs. Shear studs are

unforgiving in that they can protrude through the top of the slab when too little camber is relieved under the


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actual load. Alternatively, allow sufficient slab thickness to account for reduced actual deflection. For further

information, see Ricker.

Another thing to keep in mind: the minimum length of a beam that is to be cambered is about 25 ft. Why?

Because the fabrication jig that is used to camber beams is usually configured with pivot restraints that hold

the beam from 18 ft to 20 ft apart. To make sure there is adequate beam extending beyond this point to resist

the concentrated force from the cambering operation, a 25 ft beam is generally required.

Favor the use of partially composite action in beam design.

Although shear stud installation costs vary widely by region, on average, one installed shear stud equates to 10

lb of steel. Fully composite designs are not usually the most economical because the average weight savings

per stud is less than 10 lb. Sometimes, the average weight savings per stud for 50 to 75 percent composite

beams can exceed the point of equivalency. In some cases, non-composite construction can be most

economical. For further information, see Lorenz and Ruddy. A caveat: make sure that the beam in a composite

design is adequate to carry the weight of the wet concrete.

When composite construction is specified, the size, spacing, quantity and pattern of placement of shear stud

connectors should be specified. It should also be compatible with the type and orientation of the steel deck

used.

When evaluating the relative economy of composite construction, keep in mind that most shear stud

connector installers charge a minimum daily fee. So, unless there are enough shear stud connectors on a job to

warrant at least a day’s work, it may be more economical to specify a heavier non-composite beam.

Shear stud connectors should be field installed, not shop installed. Otherwise, they are a tripping hazard for

the erector's personnel on the walking surface of steel beams.

Consider cantilevered construction for roofs and one-story structures.

Cantilevered construction was invented primarily to reduce the weight of steel required to frame a roof.

Although today we are less concerned with weight savings than labor savings, cantilevered construction may

still be a good option. Why? Because the associated connections of the members are generally simple to

fabricate and fast and safe to erect. So cantilevered construction is still very much a potential way to save

money.

Use rolled-beam framing in areas that will support mechanical equipment.

It always happens. The structural design is performed based upon a preliminary estimate of the loads from the

mechanical systems and units. Later, the mechanical equipment is changed and the loads go up -- way up --

sometimes after construction has begun. Rolled-beam framing offers much greater flexibility than other

alternatives, such as steel-joist framing, to accommodate these changing design loads.

Optimize bay sizes.

It is still a good idea to design initially for strength and deflection. Subsequently, geometry and compatibility

can be evaluated at connections, with shape selections modified as necessary. Ruddy suggested that using a

bay length of 1.25 to 1.5 times the width, a bay area of about 1000 ft2, and filler beams spanning the long

direction combine to maintain economical framing. But, …

Avoid shallow beam depths that require reinforcement or added detail material at end connections.

Detail material such as reinforcement plates at copes and haunching to accommodate deeper, special

connections is typically far more expensive than simply selecting a deeper member that can be connected

more cleanly. If the beam is changed from a W16x50 to a W18x50, the simplified connection is attained

virtually for free. And, …

Don’t change member size frequently just because a smaller or lighter shape can be used.

Try to get the usage of any given member size to a mill order quantity (approximately 20 tons). Of course

smaller quantities can be used and are commonly purchased by the fabricator from service centers (at a cost

premium), but detailing, inventory control, fabrication and erection are all simplified with repetition and


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uniformity. Keep in mind that economy is generally synonymous with the fewest number of different pieces.

This same idea applies when selecting the chords and web members in fabricated trusses. Above all, …

Select members with favorable geometries.

Watch out for connections at changes in floor elevations; a deeper girder may simplify the connection detail.

Also, watch out for W10, W8 and W6 columns, which can have narrow flanges and web depth; connecting to

either axis is constrained and difficult. It is often most helpful to make rough sketches of members to

approximate scale in their relative positions to discover geometric incompatibilities.

Use repetitive plate thicknesses throughout the various detail materials in a project.

Just like with member sizing, the use of similar plate thickness throughout the job is generally more

economical than changing thicknesses just because you can. For example, use one or two plate thicknesses for

all the column base plates. This same idea applies for other detail materials, such as transverse stiffeners and

web doubler plates.

Design floor framing to minimize the perceptibility of vibrations.

Floor vibration can be an unintended result in service when floors are designed only for strength and

deflection limit-states and an absolute-minimum-weight system is chosen. Today’s lighter construction, when

combined with the lack of damping due to partitionless open office plans and light actual floor loadings (in the

era of the nearly paperless office), has exacerbated the potential for floor vibration problems. Fortunately,

design criteria to prevent perceptible floor vibrations from occurring are available; see Murray et al.

When designing for snow-drift loading, decrease beam spacing as the framing approaches the bottom of a

parapet wall.

Reduced beam spacing allows the same deck size to be used and the same beam size to be repeated into a

parapet against which snow may drift. This is generally more economical than maintaining the same spacing

and changing the deck and beam sizes.

Minimize the need for stiffening.

When required at locations of concentrated flange forces, transverse stiffeners and web doubler plates are

labor intensive detail materials. For the sake of economy, using 50 ksi steel and/or a member with a thicker

flange or web can often eliminate them. In the latter case, consider trading some less expensive member

weight for reduced labor requirements. Always remember to reduce the panel-zone web shear force by the

magnitude of the story shear. This can often mean the difference between having to use a web doubler plate

and not. For further information, see Carter.

Economize web penetrations to minimize or eliminate stiffening.

Web penetrations in beams are often a cost-effective means of minimizing the depth of a floor system that

contains mechanical or electrical ductwork. However, if they are numerous and require stiffening, it is

probably more economical to eliminate them and pass all ductwork below the beams, if possible. Thus,

stiffening at web penetrations should be called for only if required. The use of a heavier beam, a relocated

opening, a change in the size of the opening, and the use of current design procedures can often eliminate the

need for reinforcement of beam web penetrations. If web penetrations are to be use and stiffening is required,

the most efficient and economical detail is the use of longitudinal stiffeners above and below the opening as

illustrated in Figure 2. For more information, see Darwin.


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Figure 2. Web penetration reinforcement of an I-shaped beam.

Eliminate column splices, if feasible.

On average, the labor involved in making a column splice equates to about 500 lb of steel. Consider the

elimination of a column splice if the resulting longer column shaft remains shippable and erectable. If a

column is spliced, locating the splice at 4 to 5 ft above the floor will permit the attachment of safety cables

directly to the column shaft, where needed. It will also allow the assembly of the column splice without the

need for scaffolding or other accessibility equipment. If the column splice design requires welding in order to

attain continuity, consider the use of PJP groove welds rather than CJP groove welds for economy.

Configure column base details that are erectable without the need for guying.

Use a four-rod pattern, base-plate thickness, and attachment between column and base that can withstand

gravity and wind loads during erection. At the same time, make sure the footing detail is also adequate against

overturning due to loads during erection. For further information, see Fisher and West. This reference contains

minimum column base details for various column heights and wind exposures are recommended. And, ...

Make your column base details repetitive, too.

The possibility of foundation errors will be reduced when repetitive anchor-rod and base-plate details are

used. Keep your anchor-rod spacings uniform throughout the job. Use headed rods or rods that have been

threaded with a nut at the bottom if there is any calculated uplift. Otherwise, hooked rods can also be used if

desired. Be sure to identify both the length of the shaft and the hook if so.

Allow the use of the right column-base leveling method for the job.

Three methods are commonly used to level column bases: leveling plates, leveling nuts and washers and shim

stacks and wedges. Regional practices and preferences vary. However, the following comments can be stated

in general. Leveling plates lend themselves well to small- to medium-sized column bases, say up to 24 in. Shim

stacks and wedges, if used properly, can be used on a wide variety of base sizes. Proper use means maintaining

a small aspect ratio on the shim stack, possibly tack welding the various plies of the shim stacks to prevent

relative movement and secure placement of the devices to prevent inadvertent displacement during erection

operations and when load is applied. Leveling nuts and washers lend themselves well to medium-sized base

plates, say 24 in. to 36 in., but are only practical when the four-rod pattern of anchor rods is spaced to develop

satisfactory moment resistance. Large column base plates, say over 36 in., can become so heavy that they

must be shipped independently of the columns and preset, in which case grout holes and special leveling

devices are usually required.

Don't over-specify the details of secondary members.

For example, spandrel kickers and diagonal braces can often be provided as square or bevel-cut elements that

get welded into the braced member and structural element that provides the bracing resistance with a very

simple line of fillet weld. In contrast, it is very costly to require that such secondary details be miter-cut to fit

the profile of a member or element to which it is connected and welded all-around.

Keep relieving angles in a practical size range.


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The thickness of relieving angles is normally 5/16 in. or 3/8 in. If a greater thickness is required for strength,

the basic design assumptions should be reviewed and perhaps modified. If vertical and/or horizontal

adjustment of masonry relieving angles is required, the amount of adjustment desired should be specified and

the fabricator should be allowed to select the method to achieve this adjustment, such as by slotting or

shimming. Final adjustments to locate relieving angles should be made by the mason, preferably after dead

load deflection of the spandrel member occurs.

Consider if heavy hot-rolled shapes are really necessary in lighter and miscellaneous applications.

Ordinary roof openings can usually be framed with angles rather than W-shapes or channels. As another

example, heavy rolled angles for the concrete floor slab stop (screed angles) are unnecessary if a lighter

gage-metal angle will suffice (something in the 10 to 18 gage range, depending upon slab thickness and

overhang). These lighter angles can often be supplied with the steel deck and installed with puddle welding,

simplifying the fabrication of the structural steel. Small roof openings on the order of 12 in. square or less

probably need not be framed at all unless there is a heavy suspended load, such as a leader pipe.

More Ways to Save Time and Money

40 more basic suggestions that you can use in your office practice today to work smarter, not harder -- and to

improve the economy of steel building construction.

Limit the use of different bolt grades and diameters.

It is seldom feasible to use more than one or two combinations of bolt diameters and grades on a project. The

use of different diameters for different grades simplifies the quality assurance task of ensuring that each

strength grade was used in the proper location. It also allows more shop efficiency in the drilling or punching

operations.

Use ASTM A325 bolts whenever possible.

They are strong, ductile, and reliable -- the best fastener value.

Limit bolt diameter to 1 in.

Diameters of 3/4-in. and 7/8-in. are preferred and 1-in. diameter is still within the installation capability of

most equipment. Larger diameters require special equipment and increased spacings and edge distances than

are typical.

Select the right bolt hole type for the application.

In steel-to-steel structural connections, standard holes can be used in many bolted joints and are preferred in

some cases. For example, standard holes are commonly used in girder and spandrel connections to columns to

more accurately control the dimension between column centers and facilitate the plumbing process. In large

joints, particularly those in the field, the use of oversized holes or slotted holes can reduce fit-up and assembly

time and the associated costs. For further information, see the RCSC Specification.

Open holes need not be filled for structural purposes.

Sometimes, bolt holes remain in the structure without bolts in them. Whether this is because a temporary

bolted connection was made at that location, a design modification was made during construction or for other

reasons, there is no structural consequence of not having a bolt installed in the hole. Said another way, the

hole has the same effect on the member it pierces whether there is a bolt in it or not.

As another example, consider the connection between a roof purlin that runs over the top of the rafter. It

may be possible to punch two diagonal holes in the rafter and four in the purlin, which allows greater

repetition of member configuration throughout the job and reduces the chance that the holes may get punched

opposite to what is desired, particularly if "opposite hand" members are involved.

Don't confuse the requirements for bolts and bolt holes in steel-to-steel structural connection with those for


10 of 25 1/7/2014 6:54 PM

anchor rods and anchor-rod holes.

There are many differences between steel-to-steel structural connections and steel-to-concrete anchorage

applications, including the following important ones:

While ASTM A325 and A490 bolts are most commonly used in steel-to-steel structural connections,

they are not appropriate for use in steel-to-concrete anchorage applications. For anchor rods, ASTM

F1554 is a new umbrella specification for headed, threaded/nutted and hooked anchor rods (they still

call them anchor bolts) in three material strengths: 36 ksi, 55 ksi and 105 ksi. Of course, you can also

use ASTM A36, A572, A449 (strength equivalent of A325 in rod), A354 (strength equivalent of A490

in rod). For more information, see Part 3 of Carter.

The hole sizes that are permitted in base plates and similar devices for anchor rods are larger for any

given diameter to account for the larger placement tolerances on anchor rods.

Installation is totally different. That is, pretension is sometimes specified for steel-to-steel structural

connections, but not normally for steel-to-concrete anchorage applications.

Washer requirements are also different. Anchor rods generally require thicker, larger washers that are

often made out of plate stock.

Use snug-tightened installation whenever possible.

Snug-tightened installation requirements (the full effort of an ironworker with an ordinary spud wrench that

brings the connected plies, but without any specific level of clamping force between them) recognize that most

bolts in shear need not be pretensioned; for further information, see the RCSC Specification. The maximum

shear strength per bolt is realized while the related installation and inspection costs are minimized. As given in

the AISC Specification, the cases that must be pretensioned include: tall building column splices, connections

that brace tall building columns, some connections in crane buildings, bolts subject to direct tension (AISC and

RCSC are currently considering a relaxation of this requirement for ASTM A325 bolts in non-fatigue and

non-impact applications), connections subject to impact or significant load reversal, and slip-critical

connections.

Permit the use of any of the four approved methods for pretensioning high-strength bolts, when

pretensioning is necessary.

RCSC provides four approved installation methods for high-strength bolts that must be pretensioned: the

turn-of-nut method, the calibrated wrench method, the twist-off-type tension-control bolt method and the

direct-tension-indicator method. Different fabricators and erectors have different preferences, which mostly

center on the installation cost that is associated with their use of each of these methods. When properly used

in accordance with RCSC requirements, these methods all provide acceptable results. Therefore, it is in the

interest of economy to allow the installer the flexibility to choose their preferred method and properly use it.

For more information, see the RCSC Specification.

Minimize the use of slip-critical connections.

Most connections with three or more bolts have normal misalignments that would cause some of the bolts to

be in bearing initially. Furthermore, the normal methods of erection usually cause bolts to slip into bearing

during erection. Additional slip beyond this point is usually negligible. Special faying surface requirements add

cost due to required masking or use of a special paint system. Furthermore, additional installation and

inspection requirements add cost. Therefore, generally limit usage of slip-critical connections to those cases

required by AISC/RCSC; these include: connections with oversized holes or slotted holes not perpendicular to

load; end connections of built-up members, and connections that share load between bolts and welds. For

more information, see the AISC Specification and the RCSC Specification.

Follow RCSC Specification requirements for the use of washers.

The use of hardened washers is often unnecessary; see the RCSC Specification for when they are necessary.

However, some fabricators and erectors elect to use a hardened washer under the turned element to prevent

galling of the connected material, which would otherwise increase wear and tear on the installation equipment.


11 of 25 1/7/2014 6:54 PM

Lock washers are not intended for use with ASTM A325 or A490 bolts, nor should they be specified or used.

Are your bolt threads automatically excluded from the shear planes of the joint?

For ASTM A325 and A490 bolts, a 3/8-in.-thick ply adjacent to the nut will exclude the threads in all cases

when the bolt diameter is 3/4 in. or 7/8 in. A 1/2-in.-thick ply adjacent to the nut will exclude the threads in all

cases when the bolt diameter is 1 in. or 1 1/8 in. Use a washer under the nut and you can reduce these

minimum ply thicknesses by 1/8 in. Depending upon the combination of grip, number of washers and bolt

length, a lesser ply thickness may also work with the threads-excluded condition. For further information,

refer to Carter.

Take the easy way out on prying action in bolted joints.

Prying action, the additional tension that results in a bolted joint due to deformation of the connected parts,

generally requires detailed calculations to determine the combination of bolt diameter, gage and fitting

thickness that is required to transmit the design forces. Simplify the whole process as follows. As a first step,

calculate the fitting thickness that is required to reduce the prying action to an insignificant (i.e., negligible)

amount. If this thickness is present or can be provided, there is no further need for prying action calculations.

If this thickness is excessive or cannot be provided, then use the more detailed calculation approach that

matches the proper arrangement of bolt diameter, gage and fitting thickness to transmit the loads with prying

action.

Configure welded joints to minimize the weld metal volume.

Each pound of weld metal has a deposition time (and labor cost) associated with it. Therefore, the most

economical welded joint will generally result when weld metal volume is minimized. Furthermore, reducing

the weld metal volume reduces the heat input and the resulting shrinkage and distortion. Minimizing the weld

metal also minimizes the potential for weld defects.

Favor fillet welds over groove welds.

Fillet welds generally require less weld metal than groove welds. Additionally, the use of fillet welds virtually

eliminates base metal preparation, which is labor intensive.

Configure fillet weld length to reduce weld size.

Compare a 1/4-in. fillet weld 12-in. long with a 1/2-in. fillet weld 6-in. long. These welds have equal strength,

but the latter weld requires twice as much weld metal volume and cost. With due consideration of the

implications of increased weld length on the size of connecting elements, such as gusset plates, the best

balance can be found.

Keep fillet weld sizes at or below 5/16-in., when possible.

Per AWS D1.1, this weld size can be deposited in one pass with the shielded metal arc welding (SMAW)

process in the horizontal and flat positions. Larger weld sizes will require multiple passes.

Don't always weld on both sides of a piece just because you can.

There are many applications where it may be possible to weld on one side of a joint only. For example, the

attachment of a column base plate to a column can in many cases be made with fillet welds on one side of

each flange and the web. This same idea is also sometimes possible with transverse stiffeners, bearing

stiffeners and other similar elements.

Use intermittent fillet welds when possible.

Under typical loading, intermittent fillet welds can often be specified and the weld metal volume can be

reduced accordingly. However, in applications that involve fatigue, intermittent fillet welds are not permitted.

Favor the horizontal and flat positions.

These positions use the base metal and gravity to hold the molten weld pool in place, allowing easier welding


12 of 25 1/7/2014 6:54 PM

with a faster deposition rate and generally higher weld quality.

Recognize the increased strength in transversely loaded fillet welds.

It has long been known that a fillet weld loaded transversely is up to 50 percent stronger than the same weld

loaded longitudinally. AISC Specification Appendix J2.4 provides a means to take advantage of this strength

increase [Lesik and Kennedy]. As a result, values in the eccentrically loaded weld group tables in the current

AISC Manual are typically from ten to 30 percent higher than in the previous edition. This is particularly

useful for transverse stiffener end welds, which are purely transversely loaded and qualify for a full 50 percent

increase in shear strength.

Avoid the use of the weld-all-around symbol.

This welding is excessive in most cases. It may even be wrong in some. Welding all around may violate the

AISC Specification requirement that welds on opposite sides of a common plane be interrupted at the corner

(i.e., when the weld would have to wrap around the corner at an overlap). Also, the weld-all-around symbol

should not be used if the entire perimeter of the weld cannot be reached.

Consider the AWS acceptance criteria when an undersized weld is discovered.

AWS D1.1 allows a 1/16-in. undersize to remain if it occupies less than ten percent of the weld length. This

recognizes that an attempted repair may create a worse condition than the undersized weld.

Avoid seal welding unless it is required.

Seal welding is generally unnecessary, unless the joint is required to be air-tight or water-tight.

For groove welds, favor partial-joint-penetration (PJP) over complete-joint-penetration (CJP).

PJP groove welds generally require less base metal preparation and weld metal. They also reduce heat input,

shrinkage, and distortion. It is sometimes possible to increase plate thickness and use a PJP groove weld

instead of a CJP groove weld.

Select a groove-welded joint with a preparation that minimizes weld metal volume.

Depending upon thickness, a particular combination of root opening and bevel angle will minimize weld metal

volume. The combination that requires the least amount of weld metal should be selected. Also, consider

double-sided preparation. In some cases, the additional labor to prepare the surface can be offset by savings in

weld metal volume (and labor).

Watch out for weld details that will likely cause distortion as the welds cool and shrink.

Welding (and in many cases, flame cutting) causes distortion because the heated regions are restrained by the

rest of the steel as they cool and contract. Under extreme circumstances, a structural member could be

distorted so severely that straightening of the member would be required, particularly in an application that

involves architecturally exposed structural steel, with the resultant increase in cost. The potential for distortion

frequently can be reduced through proper selection of the connection configuration and joint details. The

fabricator is the best source of guidance and advice for avoiding potentially troublesome details.

Know what to look for when monitoring interpass temperatures in welded joints.

For a given level of heat input, the interpass temperature in a weldment is largely dependent upon the cross-

sectional area of the element(s) being welded. The larger the area, the faster the heat will be drawn from the

weldment. As a rule of thumb, if the cross-sectional area of the weld is equal to or greater than 40 in.2, the

minimum interpass temperature should be monitored. Conversely, if the cross-sectional area of the weld is

equal to or less than 20 in.2, the maximum interpass temperature should be monitored.

Avoid welding on galvanized surfaces.

When at all possible, avoid design situations that will require welding on galvanized surfaces, particularly in

the shop. Special ventilation must be provided in the shop to exhaust the toxic fumes that are produced.


13 of 25 1/7/2014 6:54 PM

Additionally, the galvanizing must commonly be removed by grinding in the area to be welded. This requires

the subsequent touching up with cold galvanizing compound after welding and cleaning. All of these

operations add cost.

Consider the fabricator's and erector's suggestions regarding connections.

To a large extent, the economy of a structural steel frame depends upon the difficulty involved in the

fabrication and erection, which is a direct function of the connections. The fabricator and erector are normally

in the best position to identify and evaluate all the criteria that must be considered when selecting and

detailing the optimum connection, including such non-structural considerations as equipment limitations,

personnel capabilities, season of erection, weight, length limitations and width limitations. The fabricator will

also know when variations in bolt diameters and holes sizes, broken gages, combination of bolting and welding

on the same shipping piece will incur excessive and costly material handling requirements in the shop.

Design connections for actual forces.

Or at least do not overspecify the design criteria. In U.S. practice, the Engineer of Record sometimes specifies

standard reactions for use by the connection designer. These standard reactions can sometimes be quite

conservative; look at the extreme example illustrated in Figure 3.. However, design for the actual forces

generally allows more widespread use of typical connections, which improves economy. Axial forces, shears,

moments and other forces should be shown as applicable so that proper connections can be made and costly

overdesign, as well as dangerous underdesign, can be avoided. This applies to shear connections, moment

connections, bracing connections, column splices -- all connections! The actual reactions are quite important

for the proper design of end connections for beams in composite construction.

Figure 3. Extreme example of what arbitrary connection criteria can mean.

Use one-sided shear connections, when possible.

One-sided connections such as single-plates and single-angles have well-defined performance, are economical

to fabricate and are safe to erect in virtually all configurations. When combined with reasonable end-reaction

requirements, one-sided connections can be used quite extensively to simplify construction. Sometimes,

however, end reactions are large enough to preclude their use because of the strength limitations of such

connections.

Avoid through-plates on HSS columns; use single-plate shear connections whenever possible.

A single-plate connection can be welded directly to the column face in all cases where punching shear does

not control and the HSS is not a slender-element cross-section. See the AISC HSS Connections Manual.

Consider partially restrained (PR) moment connections.

PR moment connections can provide adequate strength and stiffness for many buildings, particularly those

with long frame lines where many connections of reduced stiffness can be mobilized. PR connections can be

configured to minimize field welding, simplify the connection details, and speed erection. Engineers can obtain

guidance on this subject from a number of sources, including Leon et al.


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Design columns to eliminate web doubler plates (especially) and transverse stiffeners (when possible) at

moment connections.

The elimination of labor-intensive items such as web doubler plates and stiffeners is a boon to economy. One

fillet-welded doubler plate can generally be equated to about 300 lb of steel; one pair of fillet welded

stiffeners can generally be equated to about 200 lb of steel. Additionally, their elimination simplifies weak-axis

framing. For further information, see Carter.

Recent Developments

The Economy Equation, Ways to Save Time and Money and More Ways to Save Time and Money are almost

entirely upon currently established design and construction practices. The following recent developments also

have great potential to improve the economy of steel building construction.

Snug-tight bolts in tension applications

The economic benefits of snug-tight bolts can currently be realized in the majority of shear/bearing

connections in building structures. However, the current RCSC and AISC Specifications require all bolts in

tension applications to be pretensioned. Murray and Johnson showed that the pretension is not critical to

performance for ASTM A325 bolts in applications involving tension but not fatigue or impact. Accordingly, it

is anticipated that both the RCSC and AISC Specifications will permit snug-tight ASTM A325 bolts in such

applications. Thus, the potential for economy in bolted connections will be increased due to the relaxation of

the installation and inspection requirements.

Use of one-sided shear connections

The recent development of detailed design procedures for one-sided connections, coupled with erection safety

considerations, has led to a more widespread usage of one-sided shear connections, such as single-plate,

single-angle and tee shear connections.

Extended end-plate moment connections for seismic loading

With all welding done in the shop with better quality control and lower cost and only bolting in the field, the

advantages of extended end-plate moment connections have long been known. However, these connections

have historically been limited to use in static loading applications and non- and low-seismic applications.

Meng and Murray showed that, if the design procedure is modified slightly to account for the effects of

prying action and weld access holes are not used to make the beam-flange-to-end-plate welds, extended

end-plate moment connections are suitable for high-seismic loading. The use of weld access holes causes an

increase in the strain rate in the beam flange under load and results in a premature flange fracture.

At the same time, improvements in cutting and drilling equipment have made large moment end-plate

connections an increasingly viable alternative for moment resisting frames. Current cutting methods can

produce an acceptably square beam end without milling. Additionally, computer-controlled drilling equipment

can produce accurate hole placement in both the end plates and the column flanges, resulting in excellent

alignment for field assembly.

Other Seismic Moment Connections

Tremendous advancements have also been made in the inelastic deformation capabilities of welded beam-to-

column moment connections for seismic applications. At the same time, more stringent performance

requirements have been implemented in the AISC Seismic Provisions. Several connection alternatives that can

withstand inelastic rotations of at least 3 percent have been developed using reinforcement, including cover

plates, ribs and haunches [FEMA and FEMA]. Additionally, the reduced beam section (dogbone) approach

has also demonstrated excellent inelastic performance [Iwankiw, Engelhardt and Carter and Iwankiw]. At the

same time, bolted solutions, such as extended end-plates (see above), flange tees and flange plates are being

investigated further [Meng and Murray, FEMA and FEMA]. Some proprietary alternatives also have been

developed. Further information on seismic moment connections is also available in Engelhardt and Sabol.

Research is continuing in this area through the SAC Joint Venture, a partnership formed by the Structural


15 of 25 1/7/2014 6:54 PM

Engineers Association of California (SEAOC), Applied Technology Council (ATC) and California Universities

for Research in Earthquake Engineering (CUREe).

Connections Involving HSS

The AISC HSS Connections Manual has drawn together information on connections for HSS from a variety of

sources. It includes specific coverage of shear, moment, bracing, truss and other connections as they are

commonly found in tubular construction for buildings.

Framing for Reduced Floor-to-floor Heights

The primary design characteristic for some buildings is a reduced floor-to-floor height. Although this

characteristic has often led to construction in reinforced concrete in the past, steel systems are increasingly

common. Some layouts have utilized conventional framing to achieve a floor-to-floor height as small as 8 ft-8

in. with the structure well integrated into the partitions and other architectural features. Other framing systems

such as the staggered truss system [Cohen] have also been used to achieve such a reduction.

Several promising recent advancements also deserve mention. In the United Kingdom, a system called

Slimflor has been developed, wherein the infill beams are nested inside and supported by the wider bottom

flange of an asymmetrical girder [Lawson et al.]. Spans in the range of 20 to 30 ft are practical for very

shallow framing depths. Additionally, in the United States, a preliminary study by Murray on a two-way

composite floor system spanning 30 to 40 ft between similar asymmetrical girders indicates the potential for a

very competitive system. The two-way composite floor system illustrated in Figure 4 is constructed with deep

metal deck spanning one direction, a very shallow form deck spanning the other direction, and double-headed

self-tapping screws, which interconnect the two plies of metal deck and achieve composite action with the

topping concrete. For more information, see Hillman and Murray, Hillman and Murray and Hillman and

Murray.

Figure 4. A two-way steel-deck composite floor system.

Composite Trusses

Composite "joists", which are actually light trusses with composite action, have recently been introduced. This

type of construction allows for economical ultra-clear span construction, and has been used in applications

with spans from 45 to 120 ft. Ductwork and other mechanical system components can pass through the

open-web trusses, allowing for reduced building height without the need to create web penetrations as for

solid-web members. In-service measurements of floor accelerations have shown that floor vibrations are not a

significant problem in these systems. For more information, see Easterling et al., Easterling et al. and Murray.

The Uniform Force Method for Bracing Connections

Until recently, there has not been a systematic approach to the analysis and design of bracing connections,

such as the connection shown in Figure 5. There is now available just such a method which is based on

analytical and experimental results and not on the usual simple beam and strut formulas applied to various cut


16 of 25 1/7/2014 6:54 PM

sections. The method is called the Uniform Force Method and has been adopted by AISC for use in the AISC

Manual.

Figure 5. A typical bracing connection.

The admissible force distribution for this method is shown in Figures 6 and 7. The force distribution is called

admissible in the sense of the lower bound theorem of limit analysis because it satisfies equilibrium for the free

body diagrams shown in Figures 6 and 7, i.e., the gusset in Figure 6 and the beam and column and Figure 7,

with absolutely no additional forces required anywhere.

Figure 6. Force distributions in gusset plates per the Uniform Force Method.

Figure 7. Force distributions in the beam and column per the Uniform Force Method.


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The idea for the coincident system of forces shown on the gusset free body diagram of Figure 6 comes from

the work of Richard, who showed that the force resultants on the gusset edges fall within the regions shown

cross-hatched in Figure 8. Each cross-hatched region of Figure 8 contains the resultants for six cases in which

the connections of the gusset to the beam and column were varied from bolted to welded. It can be seen by

comparing Figure 8 with Figure 6 and the uniform force method captures the essence of Richard’s results.

Figure 8. Gusset edge force resultant envelopes for

45-degree working point models (adapted from Richard).

In order to validate the method, it was used to predict the failure load and controlling limit state of six

full-scale test specimens for which the actual failure load controlling limit are known. A typical test specimen

of Bjorhovde and Chakrabarti is shown in Figure 9 and that of Gross and Cheok is shown in Figure 10. Table 1

shows the limit states that have been identified for these specimens, and Table 2 shows the limit states applied

to each connection interface.

Figure 9. Test configuration used by Bjorhovde and Chakrabarti.


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Figure 10. Test configuration used by Gross and Cheok.

Table 3 shows the results of analyzing the six test specimens by the uniform force method, and comparing

the results with the actual physical test results. Predicted and actual connection failure interfaces, limit states,

and failure loads are compared. In all cases, the uniform force method gave a conservative prediction of the

connection capacity. The first three tests, those of Bjorhovde and Chakrabarti, show extremely close

correspondence between the predictions and the physical tests. The second three tests, those of Gross and

Cheok, show that the predictions of the uniform force method underestimate the test loads, that is, the uniform

force method gives a conservative prediction of capacity. There are two reasons for this.

As Gross points out, his tests include gusset buckling, and the current method for gusset buckling, treating

the gusset as a strip column, is known to be conservative because it ignores the elastic support provided by the

continuity of the gusset plate. Bjorhovde’s tests included only a brace tensile load, so this effect was not

addressed.

The second reason involves the distortion of the frame which is ignored by the uniform force method.

Gross’ tests included this but Bjorhovde’s did not. When the frame distorts, the forces on the connection

interfaces undergo a redistribution somewhat analogous to what happens during shakedown in plastic analysis.

The resulting internal force distributions are more benign than those predicted based on pure equilibrium

considerations and result in increased load carrying capacity. A discussion of this phenomenon is given by

Thornton and Kane.

Table 1. Limit-states identification for bracing connections

Limit state Number

Bolt shear fracture 1

Bolt shear/tension fracture 2

Whitmore yielding 3

Whitmore buckling 4

Tear-out fracture 5

Bearing 6

Gross section yielding 7

Net section fracture 8

Fillet weld fracture 9


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Beam web yielding (beyond

k-distance)10

Bending yielding (including

prying action)11

Bending fracture (including

prying action)12

Table 2. Limit-states considered for each interface of bracing

connections

Connection

interface (note 1)

Connection

element

Limit states (note

2)

Brace-to-gusset (A)

Bolts to gusset 1

Gusset 3, 4, 5, 6

Bolts to brace 1

Brace 5, 6, 7, 8

Splice plates for

WT's5, 6, 7, 8

Gusset-to-beam (B)

Gusset 7

Fillet weld 9

Beam web 10

Gusset-to-column

(C)

Bolts to gusset 1

Fillet weld to gusset 9

Gusset 6, 7, 8

Bolts to column 2

Clip angles 6, 7, 8, 11, 12

Column 6, 11, 12

Beam-to-column (D)

Bolts to beam web 1

Fillet weld to beam

web9

Beam web 6, 7, 8

Bolts to column 2

Clip angles 6, 7, 8, 11, 12

Column 6, 11, 12

(1) See Figure 9 for identification letters of the connection

interfaces.

(2) See Table 1 for identification number of limit state.

Table 3. Comparison of Uniform Force method predicted results with test results.

Test

specimen

Predicted results (note 1) Test results (note 1) Test

divided

by

predicted

A,

kips

B,

kips

C,

kips

D,

kips

Strength,

kips

Controlling

interface:

Strength,

kips

Controlling

interface:

Bjorhovde/

Chakrabarti

30º

142

(3,5)

184

(7)

216

(5)

152

(12)

142

(3,5)A

143

(5)A 1.01


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Bjorhovde/

Chakrabarti

45º

142

(3,5)

182

(7)

164

(5)

210

(12)

142

(3,5)A

148

(5)A 1.04

Bjorhovde/

Chakrabarti

60º

142

(3,5)

169

(7)

155

(5)

342

(12)

142

(3,5)A

158

(5)C 1.11

Gross/Cheok

#1

73

(4)

212

(7)

67

(12)

149

(9)

67

(12)C

116

(4)A 1.73

Gross/Cheok

#2

78

(4)

77

(7)

143

(7)

--

(note

2)

77

(7)B

138

(4)A 1.79

Gross/Cheok

#3

84

(4)

94

(7)

171

(7)

--

(note

2)

84

(4)A

125

(5)A 1.49

(1) Numbers in parentheses below the strengths in this table are limit-state numbers as given

in Table 1.

(2) No limit. This part of the connection does not carry any of the brace load.

Future Trends and Needs

In the opinion of the authors, Charles J. Carter, Thomas M. Murray and William A. Thornton, the following

trends and needs are critical to the continued advancement of the state-of-the-art in steel design and

construction.

Simplification of Analysis Requirements

Although consideration of second-order effects in the structural analysis is a Specification requirement in all

cases, the actual impact of second-order effects is negligible in many practical cases. Most building structures

in the United States are not taller than four stories and the actual increase in moments due to deformations of

the frame is in many practical cases negligible. At the same time, there are also cases where the consideration

of second-order effects is quite important. Some work has already been done [ASCE], but additional

simplification of analysis requirements, particularly the Specification requirements, is needed.

Composite Framing Systems

It would be advantageous for the steel design community and construction industry to innovate with practical

systems that marry structural steel and reinforced concrete for each of their benefits. For example, a system

utilizing composite connections between steel floor framing and concrete-encased steel columns (also known

as steel-reinforced-concrete) or reinforced concrete columns offers a decided advantage because of all the

deflection problems and structural inefficiencies associated with concrete-slab floor-framing. Such hybrid

systems would provide increased column spacing flexibility and improved control of floor deflection and

vibration characteristics. Additionally, for seismic design applications, these systems would provide increased

structural damping and improved structural ductility.

Development of More Advanced Column Stiffening Design Procedures

Current Specification provisions for the design of transverse stiffeners and web doubler plates do not consider

that the presence of the web doubler plate may eliminate the need for the transverse stiffeners. The interaction

between these elements and effect of the presence of one on the demands on the other should be investigated.

Additionally, alternative reinforcement approaches to costly web doubler plates should also be investigated.

Development of a Design Approach for Fire

The current and conventional approach to treatment of fire in a steel structure is to follow prescriptive

guidelines for the protection of the structural system against the heat generated by the fire. The actual

demands on the structural system in a fire are known to be quite conservative in relation to what the

prescriptive system provides protection against. Because fire protection is a significant cost associated with


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steel construction, the progression toward a less prescriptive system based more on design and engineering is

unavoidable. Recent testing in the United Kingdom and elsewhere as reported in Dowling and Burgan

emphasizes that a shift toward engineered fire protection would have a significant and beneficial impact on the

economy of structural steel buildings.

References

Note: many of the following references are listed in our feature Technical Bibliography.

American Institute of Steel Construction, Code of Standard Practice for Steel Buildings and Bridges, June

10, 1992, AISC, Chicago, IL, USA, 1992.

American Institute of Steel Construction, Hollow Structural Sections Connections Manual, AISC, Chicago,

IL, USA, 1997.

American Institute of Steel Construction, Load and Resistance Factor Design Specification for Structural

Steel Buildings, December 1, 1993, AISC, Chicago, IL, USA, 1993.

American Institute of Steel Construction, Load and Resistance Factor Design Manual of Steel Construction,

AISC, Chicago, IL, USA, 1994.

American Institute of Steel Construction, Seismic Provisions for Structural Steel Buildings, AISC, Chicago,

IL, USA, 1997.

American Society for Testing and Materials, ASTM A6/A6M, ASTM, Conshohocken, PA, 1997.






American Society for Testing and Materials, ASTM A500, ASTM, Conshohocken, PA, 1997.




American Society for Testing and Materials, ASTM A992, ASTM, Conshohocken, PA, 1997.

American Society for Testing and Materials, ASTM F1554, ASTM, Conshohocken, PA, 1997.

American Society of Civil Engineers, Effective Length and Notional Load Approaches for Assessing Frame

Stability: Implications for American Steel Design, ASCE, Reston, VA, USA, 1997.

American Welding Society, AWS D1.1 Structural Welding Code—Steel, AWS, Miami, FL, USA, 1998.

Bjorhovde, Reidar and Chakrabarti, S.K., "Tests of Full Size Gusset Plate Connections," ASCE Journal of


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