Design Procedures for Extended Shear Tabs

DesignProcedures for

Extended ShearTabs

Al Ghorbanpoor

Donald R. Sherman

Author

Al Ghorbanpoor is professor ofstructural engineering and

chairman of the Department ofCivil Engineering and Mechanicsat the University of Wisconsin-Milwaukee. He received hisPh.D. degree from the Universityof Maryland in 1985 and is aRegistered Professional Engineer.Prior to joining the University ofWisconsin-Milwaukee in 1986, heworked for seven years as a pro-ject engineer and later as chiefstructural engineer in constructionand in consulting engineeringfirms in the Washington D.C.area.

Dr. Ghorbanpoor's teachingand research include fatigue andfracture, steel connections, non-destructive testing, analysis anddesign of structures, and finiteelements. He has been a princi-pal investigator in over forty spon-sored research studies and haspublished frequently on relatedtopics. He is a Fellow of theAmerican Society of CivilEngineers and holds membershipin a number of other professionalengineering societies. Dr.Ghorbanpoor has served as thechair and member of a number ofnational technical committees ofASCE, ACI, ASNT, TRB and PTI.

Author

Professor Sherman received hisbachelor of science and mas-

ter of science degrees in civilengineering from Case Institute ofTechnology. His Ph.D. wasobtained from the University ofIllinois in 1964.

Dr. Sherman was a facultymember at University ofWisconsin-Milwaukee for 31years. Prior to joining the facultyat the University of Wisconsin, heworked as a test engineer for thePittsburgh Testing Laboratoriesand as a design engineer forAerojet-General and EssoResearch & Engineering. Muchof his industrial experience hasbeen concerned with the design

15-1

of pressure vessels and pipelines.He has been active in researchand consulting in the design ofmetal structures, particularly tubu-lar structures and equipmentdesigned for seismic loads. Hismajor area of study is tubularstructures, and he has authoredover 40 publications on the designof tubular members and connec-tions.

Professor Sherman is a mem-ber of the AISC SpecificationCommittee and chairs the TaskCommittee on Structural Tubing.He is also the past chairman ofthe Structural Stability ResearchCouncil and has been involved inits activities for over 30 years. Heis a member of the AmericanSociety of Civil Engineers and hasbeen active in several nationaltechnical committees, including aterm as chairman of theCommittee on Metals. Dr.Sherman is a ProfessionalEngineer in Ohio and Wisconsin.

Summary

This session will focus on tworecent research projects

"Design Procedures for ExtendedShear Tab Connections" and"Tests on Simple ConnectionsIncluding Slab Effects". The firstpaper presents new design proce-dures for shear connections usingextended shear tabs, whichreduces costs by eliminating theneed for expensive coping orflange reduction of framingbeams. The second paper dis-cusses the use of composite PRconnections to resist seismicloading and to reduce the cost ofrepair or retrofit of FR steel framebuildings and presents informa-tion on the creation of designguidelines regarding the use ofsimple connections in lateral load-resisting systems.

© 2003 by American Institute of Steel Construction, Inc. All rights reserved.This publication or any part thereof must not be reproduced in any form without permission of the publisher.

DESIGN PROCEDURES FOR EXTENDED SHEAR TABS

INTRODUCTION

Extended shear tabs welded to the webs of columns or girders are economically attractive since they eliminatethe necessity of coping beams framing into a connection in order to avoid interference with supportingmember flanges. When the bolt line on the shear tab extends 3-in. beyond the flanges of the supportingmember, the detailing and tolerances of the beam are the same as typical tabs welded to a column flange.There are three possible weld configurations for connecting an extended tab to the supporting member.

1. Use only a pair of vertical fillet welds between the tab and the web of the supporting member. Thisproduces a large spacing between the weld and the bolt line and larger than typical eccentricities of theshear force. It also requires thicker tabs to avoid buckling.

2. In addition to vertical welds, add a pair of horizontal welds between the top of the tab and the top flangeof a supporting girder. There will now be the typical 3-in. spacing between the end of the horizontal weldand the bolt line. However, the large distance between the vertical weld and the bolt line still exists at thebottom of the tab. This detail can also be used with supporting columns if a stiffening or continuity plateis welded between the column flanges at the top of the tab.

3. Stiffening plates can be welded between the column flanges at both the top and bottom of the tab so thepairs of horizontal welds can be placed at both locations. This is in addition to the vertical welds to theweb. The 3-in. spacing between the horizontal welds and bolt line exists at both the top and bottom of thetab.

All of these details have been used in practice, but there are no uniform design procedures. Consequently, aprogram of 17 tests supplemented with finite element analyses was undertaken. The research identified thecritical limit states and the eccentricities of the shear force relative to the weld and bolt line. From theseresults, a design procedure is recommended.

TEST PROGRAM

There were four combinations of supporting members and beams used in the experimental program. Theseincluded extremes of beam stiffness with L/d of 10 and 23. Web stiffness of the supporting members rangedfrom h/t of 22 to 54.

Group 1- W14x53 girder with 30-ft. long W12x87 beamGroup 2- W24x55 girder with 20-ft. long W18x71 beamGroup 3- W8x31 column with 30-ft. long W12x87 beamGroup 4- W14x90 column with 20-ft. long W18x71 beam

In all tests, the column segments were 8-ft. long and the girder segments were 10-ft. long.

Connections to the W12 beam used three bolts and connections to the W18 beam used five bolts. All boltswere ¾-in. diameter A325. Both standard (STD) and short slotted (SSL) holes were used in the tabs. Boltswere installed snug tight in standard holes and fully tightened when slotted holes were used. Table 1 showsthe variables for the four tests with no horizontal welds, which are referred to as unstiffened tab tests. Onlyslotted holes were used in these tests.

15-3© 2003 by American Institute of Steel Construction, Inc. All rights reserved.

This publication or any part thereof must not be reproduced in any form without permission of the publisher.

The tab thickness was determined to meet stability limit states under the anticipated maximum shear.Considering the tab as a cantilever beam of narrow rectangular cross section under an end load applied at thecentroid of the section, the critical end load is

(1)

where: t is the tab thicknessL is the tab lengtha is taken as the spacing between the bolt line and weld

For tab proportions, the second square root term is approximately unity, and the equation can be solved for arequired thickness.

(2)

In addition, the tab thickness should meet the stability criteria in the AISC Manual (1994) for tabs.

(3)

TABLE 1 - VARIABLES FOR UNSTIFFENED TESTS

TEST

1-U2-U3-U4-U

SUPPORTMEMBER

W14x53 gird.W24x55 gird.W8x31 col.W14x90 col.

TAB t(in.)

3/83/83/81/2

WELDSIZE(in.)

5/165/165/165/16

WEBh/t

30.854.622.225.9

# ofBOLTS

3535

TABLENGTH

(in.)

915915

WELD-BOLTSPACE

(in.)

6.856.306.8610.04

Table 2 shows the variables in the remaining tests. All tabs in these tests were 1/4-in. thick and the filletwelds were 3/16-in. The four Groups of beam and support combinations had one test with standard holes andone test with short slotted holes. A few variations in the weld configurations were also tested. In Test 2-Cthe fillet weld to the column web was doubled in size and placed on only one side of the tab. In Test 3-C noweld was used at the column web and the tab was welded only to the top and bottom stiffening plates. Thestiffening plates were generally welded to column flanges but not to the column web. However, since thestiffening plates could be continuity plates for orthogonal framing, the plates were also welded to the web inTests 3-D and 3-E. Test 3-E also had an extra long tab to reflect a condition where the continuity platespacing is considerably larger than required for the 9-in. tab length.

Twisting of the tabs was evident in some of the earlier tests. Therefore, lateral bracing to prevent twist of thebeam was used in several tests. This bracing was at the location where the beam was loaded, which is close tothe connection.

15-4


In stiffened connections, the tab is supported by the horizontal weld to within 3-in. of the bolt line and thestability criteria of Equation (2) does not apply. Only Equation (3) must be satisfied.

TABLE 2 - VARIABLES FOR STIFFENED TESTS

TEST

1-A1-B2-A2-B2-C3-A3-B3-C3-D3-E4-A4-B4-C

SUPPORTMEMBER

W14x53 gird."

W24x55 gird."

W8x31 col."

W14x90 col.""

# ofBOLTS

3355533333555

HOLETYPE

STDSSLSTDSSLSTDSTDSSLSTDSTDSTDSTDSSLSTD

TABLENGTH

(in.)

99151515999919151515

WELD*CONFIG.

w-tw-tw-tw-t

w(3/8", one side)-tw-t-bw-t-bt-b

w-t-b **w-t-b **

w-t-bw-t-bw-t-b

WELD c.g.TO BOLT

SPACE(in.)6.506.505.985.985.985.915.915.915.916.238.258.258.25

BRACING

nononoyesnononononononoyesyes

* Weld Configurations: w = web; t = top; b = bottom** Stiffening plates welded to column web

TEST SETUP AND SPECIMENSThe test setups for the four Groups of beams and support members are shown in Figure 1, where two havecolumns and two have girders as the supporting member. The position of the concentrated load is such thatthe shear reaction and end rotation of each beam are the same as for a uniformly loaded beam with length of:

(4)

where: is the length of the test beam with a concentrated loadLpb is the length between the load and the far end of the beam.

The W12x87 beam used with the three-bolt connections has a length to depth ratio (Lu /d) of 23, representing aflexible beam. The stiff W18x71 beam used with the five-bolt connections has of 10.

The tests were instrumented to obtain important data. Load cells measured the applied load and the reactionat the far end of the supported beam. The difference in these readings is the shear force in the connection.

Two displacement transducers measured the displacement between the supporting member and the end of thetop flange of the supported beam. The average of the transducer readings gives the connection distortion andthe difference divided by their spacing gives the twist in the connection.

Readings from tiltmeters on the supporting member and the end of the beam give the rotation of each memberat the connection.

15-5

"

"""

Lu /d


Fig. 1 - SETUPS FOR BEAM AND SUPPORT MEMBER COMBINATIONS

Three pairs of strain gages were mounted on the top and bottom flanges of the supported beams at threelocations between the bolt line and the applied load. A pair of gages reflect the bending moment at thelocation and a linear regression analysis of the data from the three pairs of gages establishes the momentgradient. By calculating where the moment is zero, the eccentricity of the shear force from the bolt line orweld is determined.

Strain gages were mounted on both sides of the tab at the top and bottom, 1-in. from the welds. They measurethe normal strain perpendicular to the weld and indicate if the tab is bending in a twisting mode. Gages werealso mounted on the back of the web of the supporting member at the top and bottom of the tab. These gageswere perpendicular to the tab and indicate the amount that the web is being bent. A rosette gage was mountedon the tab to provide information on the general state of stress. The data from the gages are primarily used toverify the connection behavior and to provide information to calibrate a finite element model.

The connection area was whitewashed so that the initiation of yielding and slip in the connection could beobserved.

15-6


The following parameters were common to all four tests:

¾-in. A325-X bolts; 3-in. pitch and 1½-in. edge distancesshort slotted holes perpendicular to the length of the tab when usedfillet welds on both sides of the tab (except Test 2-C)E70 electrode for weldingA36 material for the tabsGr50 material for the supporting members8-ft. long columns and 10-ft. long girders with tabs at midlength

MATERIAL PROPERTIES

Tensile properties of the tabs were obtained from extra plate material supplied with the specimens. Couponscut from the supporting member webs were used to determine tension properties (based on ASTM 370). Themeasured values that are used in the calculation of limit states are given in Table 3.

TABLE 3 - MATERIAL PROPERTIES

MEMBER

UNSTIFFENED3/8-in. TAB½-in. TABW14x53 WEBW24x55 WEBW8x31 WEBW14x90 WEB

STIFFENED¼-in. TAB (W8,W24)¼-in. TAB (W14s)W14x53 WEBW24x55 WEBW8x31 WEBW14x90 WEB

THICKNESS(in.)

0.3710.5060.3700.3920.2880.468

0.2460.2470.3630.3820.2760.473

YIELD(ksi)

42.640.554.255.155.256.7

44.445.755.559.255.755.5

ULTIMATE(ksi)

66.563.670.870.175.371.7

72.369.873.873.673.670.8

% ELONGATION

343638383137

333029302830

As indicated in the Table, the ¼ in. tabs for the stiffened W14 girder and column tests were from a differentmaterial than the other stiffened tests. The values reported for the stiffened webs are an average of all themembers in a Group.

TEST RESULTS

The most important information in developing a design procedure is:

the eccentricity of the shear relative to bolt linethe ultimate loadthe failure mode

The tests were terminated when the distortions posed a stability problem to the test setup. However, asindicated in data plots, either the shear-deflection or shear-twist curves approached a level condition,indicating that failure was imminent. Although the exact failure mode is sometimes difficult to determine, ajudgement as to the primary failure and observations of secondary modes that were developing was made. In

15-7© 2003 by American Institute of Steel Construction, Inc. All rights reserved.

This publication or any part thereof must not be reproduced in any form without permission of the publisher.

all cases, there was significant and permanent shear distortion of the tab, although this did not lead to a loss ofload.

EccentricityThe shear eccentricity relative to the bolt line and welds is important in evaluating several of the limit statesapplicable to tab connections. The AISC Manual (1994) contains eccentricity equations relative to the boltline for flexible and rigid supports with standard and short slotted holes. These equations involve the numberof bolts, n, and the spacing between the welds and bolt line, a.

Rigid - Standard:

Rigid - Slotted:

Flexible - Standard:

Flexible - Slotted:

(5)

(6)

(7)

(8)

From the tests of shear tabs on HSS columns, where the wall has boundary conditions similar to a W columnweb, Sherman & Ales (1991) proposed an empirical equation that involves the wall stiffness of the supportingmember (h/t for a column web) and the stiffness of the beam, L/d, in addition to the length of the tab, L. Theeccentricity in inches from the weld is:

(9)

The eccentricity from the bolts line, is determined by subtracting from the distance a. Since thisequation is based on tests with only vertical welds, it should not be used for the current stiffened weldconfigurations. The eccentricity used for stiffened weld configurations should be to the centroid of the weldgroup as defined in the LRFD Manual (1994).

Measured eccentricities are based on the pairs of strain gages mounted on the supported beam. Although thisis an extrapolation of data and not precise, the results are consistent with the failure modes observed in thetests. Data plots show that the measured eccentricity varies with the shear. However, as the ultimate shear isapproached, the eccentricity tends toward a constant value and this is the value that is critical for evaluatingthe limit states. Table 4 compares the measured eccentricities relative to the bolt line with the variouspredictions. The a distance used in the AISC equations for the stiffened tests is taken as the distance from theweld group centroid to the bolt line. In Table 4, negative values indicate that the shear force is between theweld and bolt line, while for positive values it is outside the bolt line. Although the AISC equations giveabsolute values which are satisfactory for designing the bolts, negative values have been tabulated forcomparison with the experimental values.

e b ew

15-8

,


TABLE 4 - SHEAR ECCENTRICITIES RELATIVE TO THE BOLT LINE (in.)

TEST

UNSTIFFENED1-U, W14 gird.2-U, W24 gird.3-U, W8 col.4-U, W14 col.

STIFFENED1-A, W14 gird.1-B2-A, W24 gird.2-B2-C3-A, W8 col.3-B3-C3-D3-E4-A, W14 col.4-B4-C

EXP.

-3.2-5.8-3.3-6.5

-2.8-2.2-4.3-5.0-4.9-0.3-0.22.60.50.4-1.7-0.3-1.5

AISC

Rig-STD

-4.85-2.30-4.86-6.04

-4.50—

-1.98—

-1.98-3.91

—-3.91

44

-4.73-4.25

—-4.25

Rig-SSL

-4.85-2.97-4.86-6.71

—-4.50

—-2.65

——

-3.91————

-4.92—

Flex-STD

-6.85-6.30-6.86-10.04

-6.50—

-5.98—

-5.98-5.91

—-5.91

"

44

-8.25—

-8.25

Flex-SSL

-6.85-6.30-6.86-10.04

—-6.50

—-5.98

——

-5.91————

-8.25—

S & A

n.a.n.a.

-4.40-8.51

n.a.n.a.n.a.n.a.n.a.n.a.n.a.n.a.n.a.n.a.n.a.n.a.n.a.

None of the equations give a consistent accurate prediction of the measured eccentricity. For TEST 3-U and 4-U with column supporting members, the AISC equations for rigid supports are reasonable, as are the Sherman& Ales equation. TEST 2-U with a girder supporting member correlates better with the AISC flexible support,but this is not the case for TEST 1-U. However, the girder h/t for Test 2-U is much higher.

A similar trend is evident for the stiffened connections. The Group 2 tests with a flexible girder web are inbetter agreement with the AISC flexible support equations. The other Groups agree better with the rigidequations and the AISC values are larger than the measured, which is conservative for evaluating limit statesassociated with the bolts. There is no consistent trend in the eccentricities between STD and SSL holes. Forthe Group 1 and Group 3 connections with three bolts, the AISC equations show no difference for hole type,which agrees with the measured values. For five-bolt connections, AISC predicts the same or largereccentricities for SSL holes, but the measured values in Group 4 indicate the opposite.

Ultimate Capacity and Limit States

The ultimate shear forces for the tests, are shown in Table 5. The tests were terminated when the shear-distortion curves flattened. The flattening of the curve corresponded to considerable distortion in theconnection and further loading may have caused damage to the test setup. The failure modes have also beenidentified. To some extent, that is a judgement since several conditions usually existed simultaneously.Secondary failure modes are listed in parentheses.

Vex

15-9

,


The standard AISC limit states for single plate connections that were evaluated are taken from the AISC HSSConnections Manual (1997.) This Manual identifies the limit states, provides the equation for evaluating thecapacity and gives references for the source of the equations. These limit states are:

bolt shear using an ultimate analysis that includes eccentricitybolt bearing in the tab based on the bolt shear analysisgross shear of the tabnet shear rupture in the tab at the bolt lineblock shear rupture in the tabweld capacity based on an ultimate analysis including eccentricity (However, if the AISCrecommendation of using a weld size that is ¾ of the tab thickness is used, the weld capacity willnever control.)

Beam web limit states are not considered in these tests, since they never control.

The critical AISC capacities in Table 5 are calculated using the thickness and material strengths in Table 3and the weld to bolt line spacing in Tables 1 and 2. No resistance factors are included. The AISC capacitiesare calculated using the appropriate AISC eccentricity equation and using the measured eccentricity fromTable 4. The critical limit state is identified. For the Group 2 tests with high h/t webs, the AISC eccentricityvalues for flexible supports were used, since these correlated better with the experimental eccentricities.Eccentricities for a rigid support produced capacities that were greater than the experimental loads in theGroup 2 tests, and are not reported.

In addition to the strength limit states, there are two tab thickness checks to insure rotation and preventbuckling. The ¼-in. and 3/8-in. tabs meet both criteria, but the ½-in. tab is somewhat thicker than the 0.438-in. limit (calculated by ) for rotation capacity.db /2 + 1/16

During the tests with unstiffened connections and column supporting members, considerable distortion of thecolumn web was observed. Therefore, an additional limit state of a yield line mechanism in the column webhas been included for these tests. The equation for the moment capacity of this mechanism is taken fromAbolitz & Warner (1965.)

(10)

where: is the shear eccentricity to the weldsh is the depth of the column web (nominal h/t times t)

is the column web thicknessis the column web yield strength

V3

ew

twFywL is the length of the tab

Due to differing web boundary conditions with the girder supporting members or for stiffened connections, theyield line mechanism does not apply.

Table 5 also contains the capacity, that would be predicted by AISC if the distance between the bolt lineand welds was the standard 3-in. Although this has no meaning for the tests, it is included to indicate theamount of strength reduction associated with extended shear tabs.

15-10

,


TABLE 5 - MAXIMUM SHEAR AND LIMIT STATE CAPACITIES (KIPS)

TEST

UNSTIFFENED1-U2-U3-U4-U

STIFFENED1-A1-B2-A2-B2-C3-A3-B3-C3-D3-E4-A4-B4-C

EXPERIMENTAL

58.789.354.898.7

58.354.689.092.683.353.253.122.151.148.1103107107

FAILURE

MODE

b (d,a)f (e,a,b)

e (a)f (e,a)

c (f,b)c (f,b)c,f (b)c,f (b)c,f (b)

c,fc,fc,fc,fc,fc,fc,fc,f

CRITICAL AISC

(AISC e)

36.161.731.580.9

34.134.168.668.668.638.938.9—

38.932.486.178.786.1

LIMIT

STATE

aaaa

aaaaaaa

—aaaaa

(MEAS. e)

45.670.241.064.3

49.856.785.577.978.959.059.0—

59.059.0102102102

LIMIT

STATE

aaea

aaaaacc

—ccccc

AISC TYP.

72.614272.6144

61.061.098.398.3—59.059.0—

59.0—

98.398.398.3

a - bolt shear; b - bolt bearing; c - shear yield; d - shear rupture; e - web mechanism; f - twist

A number of observations can be made from Table 5.

1. The calculated capacities are conservative compared to the test capacities. Therefore, current designprocedures with appropriate determination of the a distance are conservative.

2. Except for Test 4-U, the calculated capacities based on measured eccentricities correlate better with testcapacities than the capacities based on AISC eccentricities. The latter always produce bolt shear as thecritical limit state.

3. In Test 3-U and the stiffened tests in Group 3 and 4, a change in the eccentricity not only affects thecalculated capacity, but may also change the governing limit state.

4. The support web mechanism must be considered for unstiffened connections to columns with high

5. For unstiffened connections, the capacity is significantly less than a typical tab of the same thickness with3-in. between the weld and bolt line. However, unstiffened tabs could still be used for small beamreactions. The reduction is not as great for stiffened connections and the reduction is less than 10%,except for the elongated tab in Test 3-E. The reductions when the support is a girder with just onehorizontal weld is similar to those for supporting columns with two horizontal welds.

6. Comparing Tests A and B in each of the Groups with stiffened connections, using snug bolts in STD holesor tightened bolts in SSL holes does not influence the capacity. No tests were conducted with snug tightbolts in SSL holes.

7. From Test 3-C, the vertical weld to the supporting web is essential. This test failed at a low load whenplastic beam mechanisms developed in both the upper and lower stiffeners. Thicker stiffeners or welding

15-11

Vex Vth Ve V3


the stiffeners to the supporting web would increase the capacity, but would probably not be as effective asincluding the vertical welds.

8. Test 2-C has a somewhat lower capacity than Test 2-A, indicating that two vertical fillet welds arepreferable to a single larger weld on just one side.

9. Since Test 3-D has a similar capacity to 3-A, welding the stiffening plates to the column web, as would bedone for continuity plates, is not a significant factor.

10. Comparing Test 2-B vs. 2-A and 4-C vs. 4-A, providing lateral bracing that prevents twist to the beamdoes not have an influence on the capacity.

BEHAVIOR

Two types of behavior that were observed in the tests deserve some comments. These are the termination oflinear load-deflection behavior and twisting of the tab. Both of these were observed well before thetermination of the tests and do not represent failure limit states.

Nonlinear Behavior

In all of the tests, there was considerable permanent shear distortion that was visually evident and wasreflected in the load-deflection curves. Table 6 presents three loads concerning linear behavior. is themeasured load where the load-deflection curve became nonlinear, is when strain gauge data indicates thatshear yielding had occurred in the tab, and is the calculated load for shear yielding through the depth of thetab. The ultimate experimental load, is included as a reference.

TABLE 6 - LOADS RELATED TO LINEARITY

TEST


STIFFENED1-A1-B2-A2-B2-C3-A3-B3-D3-E4-A4-B4-C

(kips)

48653782

424464696537383840888484

(kips)

—81

>54.8>98.7

——————————9397

(kips)

85.314285.3184

59.059.098.398.398.359.059.059.059.0102102102

(kips)

54.898.754.898.7

58.354.689.092.683.353.253.151.148.1103107107

Nonlinear behavior began at nearly the same load level in all stiffened tests in a Group. In all cases, it wasobserved well below the calculated yield load. Therefore, it should be considered a system connectionphenomenon. It is not an indication of imminent failure. The calculated yield load is generally greater thanthe ultimate test load, although shear yielding was evident in gross distortion of the tab and by flaking of theapplied whitewash. This behavior was also observed in previous tests by Sherman and Ales (1991).

15-12

V1 V2 Vy Vex

V1V2

VyVex


Twisting

Particularly in the tests with five bolts in the connection, twisting was also observed. The tab and thesupported beam web twisted and a separation between them occurred at the bottom of the tab. The separationwas such that a shim could be inserted as far as the bottom bolt. A limit state was developed for pure torsionof the rectangular tab cross section with the torsion taken as the shear force times 1/3 of the sum of the tab andbeam web thickness. Considering the shear force acting at a third of the thickness was verified by finiteelement analysis that match measured strains. Failure is defined when the maximum shear stress reaches theshear yield of the material. The resulting limit state is:

(11)

The torsion limit state, average termination of load-deflection linearity, and the average ultimateexperimental strengths are in shown in Table 7.

TABLE 7 - SHEAR AT YIELD FOR TORSION

TEST


STIFFENENEDGROUP 1GROUP 2GROUP 3GROUP 3

35.760.935.793.2

19.132.619.133.8

48653782

43653885

58.789.354.898.7

568952101

There appears to be some correlation between termination of linear behavior and the torsion limit state. Sincetorsion and shear both produce vertical shear stresses in the tab, it is logical to assume that they combine insome way to initiate yielding and produce nonlinear behavior. However, analysis combining the two stressesand setting the sum equal to the shear yield, produces shear forces that are considerably below the observednonlinear behavior. The initiation of nonlinear behavior does not represent a failure limit state since it occurswell below ultimate shear forces. However, the full shear yield limit state still correlates reasonably well withexperimental strengths in some tests.

Lateral-torsional buckling was considered as a limit state associated with twist, as it is for coped beams.(Cheng and Yura, 1985). However, since the computed critical shear for full yielding is much lower than theultimate experimental shear and the addition of torsional bracing near the connection had no influence on theultimate load, lateral-torsional buckling was not considered to be a factor. In addition, load-deflection andload-twist curves did not indicate a drop in load as is typical of stability failures. At the same time, it must berecognized that the tab support has little torsional stiffness at high loads and the beam must be prevented fromrolling as a rigid body. The history of using tabs in practice indicates that lateral bracing and typical floorframing near the connection are sufficient.

15-13

V

Vex

V1t,

Vt V1


DESIGN PROCEDURE

A design procedure has been developed based on an examination of the test results presented in this paper.First, however, the limits of applicability must be defined.

1. The number of bolts should be from two to five. Since five bolt connections were the largest that weretested, extrapolation to larger connections is outside the range of the data base, and caution is required.

2. A vertical weld to the web of the supporting member is required and all welds should be in pairs on bothsides of the tab.

3. The bolt line should be 3-in. beyond the flange tip of the supporting member.4. Either STD or SSL holes may be used. Bolts should be fully tightened in SSL holes but may be snug tight

in STD holes.5. Stiffeners between column flanges do not have to be welded to the column web, but they may be if the

stiffener also serves as a continuity plate.6. Any elongated length of the tab between the column flanges should not exceed twice the length of the tab

at the connection.7. Rigid body twist of the beam must be prevented by bracing or floor framing.

For connections that meet the limits of applicability, the following design procedure may be used.

a.) Determine the tab thickness for stability. Use the larger thickness from Equations (2) and (3) forunstiffened tabs, where V is the required shear. For stiffened tabs, only Equation (3) applies.

b.) Determine the distance a from the centroid of the weld group to the bolt line. The centroid is at the weldsfor unstiffened connections and Table 8-42 or 8-44 of the AISC Manual (1994) are convenient fordetermining the weld centroid for stiffened connections.

c.) Determine the eccentricity e of the shear force relative to the bolt line. It is permissible to use the AISCequations for eccentricity. In lieu of these, a less conservative but still satisfactory design may be obtainedusing:

e = a for of the supporting member > 35e = 0.5a for unstiffened connections withe = 0.5a for stiffened connections to girders withe = 0.25a for stiffened connections to columns with

d.) Using AISC criteria (AISC HSS Connections Manual, 1997), determine the critical limit state based onbolt shearbolt bearing in the tab90% of shear yielding on the gross area (reduce an additional 10% for elongated tabs on columns)shear rupture on the net sectionblock shear rupture

e.) For unstiffened connections to column webs, also determine the yield line mechanism by Equation (10)f.) Use a weld size equal to ¾ of the plate thickness.g.) Examine the beam web for bolt bearing.

Table 8 compares the design strengths, determined by the proposed procedure with the unstiffened and theaverage of the A and B stiffened experimental strengths for each Group.

Vn

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,


TABLE 8 - COMPARISONS OF DESIGN SHEAR STRENGTHS (kips)

GROUPUNSTIFFENED

GROUP 1GROUP 2GROUP 3GROUP 4

STIFFENEDGROUP 1GROUP 2GROUP 3GROUP 4

44664378

45695391

59895599

569153105

The design strengths are conservative with respect to the experimental capacities and are between the

Vn Vex

capacities calculated using AISC and measured eccentricities in Table 5.

CONCLUSIONA design procedure for extended shear tabs has been developed. It is based on a series of 17 tests thatincluded several variations in the weld configuration to the supporting member. The procedure is amodification of existing AISC criteria for shear tabs. A yield mechanism of the web of a supporting column isa new limit state that was identified. More study is required to determine if the procedure can be used forconnections with more than five bolts

REFERENCES

Abolitz, A.L. and Warner, M.E., 1965, "Bending Under Seated End Connections," Engineering Journal,AISC, Vol. 2, No. 1, pp.1-5.

American Institute of Steel Construction, 1994, Manual of Steel Construction LRFD, Vol. II, Connections,2nd ed., AISC, Chicago, IL.

American Institute of Steel Construction, 1997, HSS Connections Manual, AISC, Chicago, IL.

Cheng, J.J.R., Yura, J.A., and Johnson, C.P., 1988, "Lateral Buckling of Coped Steel Beams," Journal ofStructural Engineering, Vol. 114, No.1, January, ASCE, New York, NY.

Sherman, D.R. and Ales, J.M., 1991, "The Design of Shear Tabs With Tubular Columns," Proceedings of the1991 National Steel Construction Conference, pp. 23.1-24.14, AISC, Chicago, IL.

ACKNOWLEDGEMENTS

The research was supported by AISC. Tests specimens were fabricated and supplied by Ace Iron & Steel Co.of Milwaukee, WI. The testing and data evaluation was conducted by a very capable graduate student, StevenRech.

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Design Procedures for Extended Shear Tabs

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