David Dieter MS Thesis 2002

EXPERIMENTAL AND ANALYTICAL STUDY OF

CONCRETE BRIDGE DECKS CONSTRUCTED WITH FRP

STAY-IN-PLACE FORMS AND FRP GRID REINFORCING

by

David A. Dieter

A thesis submitted in partial fulfillment of

the requirements for the degree of

Masters of Science

(Civil Engineering)

at the

UNIVERSITY OF WISCONSIN – MADISON

2002

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ABSTRACT

This thesis describes laboratory testing, test results, and research to develop selected design

recommendations for construction of concrete bridge decks using only a Fiber Reinforced

Polymers (FRP) reinforcement system. The use of FRP reinforcing is being pursued to

increase the durability of bridge decks. The unique aspect of the new reinforcement system

is that it will use FRP stay-in-place formwork to serve as formwork and as the bottom

transverse reinforcement of the bridge deck. In addition, a bi-directional prefabricated FRP

grid will be used for the top layer of concrete reinforcing. Full-scale prototype laboratory

testing on FRP reinforced concrete slabs and beams was conducted to compare results of

experimental load distribution widths to American Association State Highway and

Transportation Officials (AASHTO) design equations used for steel reinforcement. In

addition, testing provided performance criteria for the reinforcement system on strength

factors of safety, failure modes and fatigue. Using experimental results for verification, an

analytical model of a prototype slab/girder bridge was built to determine the negative

moment distribution width of the reinforcement system. Test results indicate AASHTO’s

effective distribution width equation for steel reinforcement is applicable. In addition, design

recommendations are given for punching shear capacity based on American Concrete

Institute (ACI) 318 Building Code Requirements. Analytical modeling indicates that the

FRP reinforcement system is adequate for use in the prototype bridge, especially from a

strength perspective.

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ACKNOWLEDGEMENTS

I would like to thank Stan Woods and Gerry Anderson from the Wisconsin Department of

Transportation for their cooperation. Without their help and proactive contributions this

project would of never materialized. I would also like to recognize Tom Strock from the

Federal Highway Administration for his constructive ideas. Also a key to this project was

Bernie Gallagher of Alfred Benesch and Company. Bernie’s exhibited innovative ideas and

endless enthusiasm during our brainstorming research and design meetings. On behalf of the

University of Wisconsin research team, I would like to thank Innovative Bridge Research and

Construction (IBRC) for the funding for which was the fuel that drove this research.

I like to extend my gratitude to Bill Lang and John Dreger at UW Structures and Materials

Testing Laboratory. Their expertise and willingness to help ensured this project was carried

out expeditiously. UW undergraduates Brian Beaulieu, Eric Helmueller, Wyatt Henderson

and Beau Sanders provided essential help during the construction, setup and testing of the

research specimens.

Finally, I would like to extend my deepest appreciation and admiration to Dr. Lawrence

Bank, Professor Michael G. Oliva, and Professor Jeffrey Russell for their effort, guidance

and understanding. I hold your wisdom, support and timely encouragement as a key for my

successful completion of this thesis. I cannot express enough about the importance of the

solid education provided by these three gentlemen.

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TABLE OF CONTENTS

ABSTRACT...............................................................................................................................i

ACKNOWLEDGMENTS .......................................................................................................ii

TABLE OF CONTENTS .......................................................................................................iii

LIST OF FIGURES ...............................................................................................................vii

LIST OF TABLES ..................................................................................................................xi

1 INTRODUCTION ............................................................................................................1

1.1 INTRODUCTION................................................................................................1

1.2 FRP REINFORCEMENT SYSTEM AND DESIGN METHODOLOGY .................3

1.3 OBJECTIVES......................................................................................................7

1.4 SCOPE................................................................................................................8

2 BACKGROUND AND LITERATURE REVIEW ........................................................10

2.1 INTRODUCTION..............................................................................................10

2.2 BRIDGE DECKS CONSTRUCTED FROM FRP COMPOSITES.........................10

2.3 CONCRETE BRIDGE DECKS REINFORCED WITH FRP BARS......................12

2.4 CONCRETE BRIDGE DECKS AND CONCRETE SLABS REINFORCED WITH

FRP GRIDS.......................................................................................................14

2.5 USE OF STEEL SIP FORMS WITH REINFORCED CONCRETE

BRIDGE DECKS...............................................................................................16

2.6 CONCRETE BRIDGE DECKS AND CONCRETE DECK PANELS

REINFORCED WITH FRP SIP FORMS.............................................................17

3 PREPERATION OF TEST SPECIMENS......................................................................19

3.1 INTRODUCTION..............................................................................................19

3.2 TEST SPECIMENS............................................................................................19

3.3 PRECAST FACILITY........................................................................................21

3.4 ASSEMBLY METHODOLOGY ........................................................................21

3.5 CONCRETE PLACEMENT AND SPECFICATIONS.........................................25

3.6 TRANSPORTATION.........................................................................................27

3.7 EVALUATION..................................................................................................27

3.8 CONSTRUCTION OF AN ADDITIONAL CONCRETE DECK PANEL..............28

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3.9 CONCRETE QUALITY CONTROL...................................................................29

4 DESCRIPTION OF TEST METHOD............................................................................31

4.1 INTRODUCTION..............................................................................................31

4.2 CONCRETE DECK PANEL TEST.....................................................................32

4.2.1 CONCRETE DECK PANEL TEST STATION .................................................33

4.2.2 DATA ACQUISITION AND INSTRUMENTATION.........................................35

4.2.3 ISTRUMENTATION APPLIED AND DATA RECORDING .............................41

4.2.4 DESCRIPTION OF SPECIMEN LOADING...................................................46

4.3 POSITIVE MOMENT BEAM TEST...................................................................48

4.3.1 POSITIVE MOMENT BEAM TEST STATION................................................48


4.3.3 INSTRUMENTATION APPLIED AND DATA RECORDING...........................51


4.4 NEGATIVE MOMENT BEAM TEST ................................................................55

4.4.1 NEGATIVE MOMENT BEAM TEST STATION...............................................56




4.5 ACCELERATED FATIGUE BEAM TEST.........................................................62

4.5.1 ACCELERATED FATIGUE BEAM TEST STATION.......................................63




4.6 AN ADDITIONAL CONCRETE DECK PANEL TEST.......................................65

4.6.1 CONCRETE DECK PANEL TEST STATION .................................................66




5 TEST RESULTS.............................................................................................................70

5.1 INTRODUCTION..............................................................................................70

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5.2 CONCRETE DECK PANEL TEST.....................................................................70

5.2.1 PANEL A, 11’-6” SPAN................................................................................71

5.2.2 PANEL B, 9’-10” SPAN................................................................................79

5.2.3 PANEL C, 8’-0” SPAN.................................................................................87

5.3 POSITIVE MOMENT BEAM TEST...................................................................94

5.3.1 BEAM D1....................................................................................................95

5.3.2 BEAM D2....................................................................................................99

5.3.3 BEAM D3..................................................................................................104

5.4 NEGATIVE MOMENT BEAM TEST ..............................................................110

5.4.1 BEAM E1...................................................................................................110

5.4.2 BEAM E3...................................................................................................117

5.4.3 BEAM E5...................................................................................................121

5.5 ADDITIONAL CONCRETE DECK PANEL TEST, PANEL C2........................124

5.6 ACCELERATED FATIGUE BEAM TEST.......................................................133

5.7 CONCRETE CYLINDER STRENGTHS ..........................................................139

6 DATA ANALYSIS OF TEST RESULTS....................................................................141

6.1 INTRODUCTION............................................................................................141

6.2 CONCRETE DECK PANELS ..........................................................................142

6.2.1 POSITIVE MOMENT DISTRIBUTION WIDTH ...........................................142

6.2.2 PUNCHING SHEAR CAPACITY.................................................................148

6.2.3 EVALUATION OF ULTIMATE CAPACITY AND EFFECTIVE STIFFNESS ..152

6.3 POSITIVE MOMENT BEAMS ........................................................................153

6.3.1 POSITIVE MOMENT CURVATURE RELATIONSHIP .................................154


6.4 NEGATIVE MOMENT BEAMS......................................................................156

6.4.1 NEGATIVE MOMENT CURVATURE RELATIONSHIP................................157


6.5 ACCELERATED FATIGUE BEAM.................................................................161

6.6 PERFORMANCE OF TEST SPECIMENS BASED ON STIFFNESS .................162

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7 ANALYTICAL MODEL TO DETERMINE NEGATIVE MOMENT

DISTRIBUTION WIDTH ...........................................................................................166

7.1 INTRODUCTION............................................................................................166

7.2 OVERALL DESCRIPTION OF THE METHODOLOGY ..................................167

7.3 DEFINING THE SECTION PROPERTIES.......................................................168

7.4 FINITE ELEMENT MODELING.....................................................................170

7.4.1 VERIFYING EXPERIMENTAL RESULTS, MODELING PANEL C2..............170

7.4.2 MODELING THE PROTOTYPE BRIDGE DECK ........................................174

8 CONCLUSIONS........................................................................................................180

8.1 INTRODUCTION............................................................................................180

8.2 OBSERVATIONS AND CONCLUSIONS........................................................181

8.2.1 SERVICEABILITY......................................................................................181

8.2.1 STRENGTH ...............................................................................................183

8.2.1 DETAILING AND CONSTRUCTION ..........................................................184

9 SUMMARY AND RECOMMENDATIONS............................................................185

9.1 DESIGN CRITERIA........................................................................................185

9.2 SRENGTH OF FRP-SIP REINFORCED DECK ................................................186

9.3 DESIGN FOR SERVICEABILITY ...................................................................187

9.4 FURTHER INVESTIGATION .........................................................................188

9.5 IMPROVING THE FRP REINFORCING..........................................................189

9.6 CONSTRUCTABILITY...................................................................................190

REFERENCES ......................................................................................................................191

APPENDIX A

APPENDIX B

APPENDIX C

APPENDIX D

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LIST OF FIGURES

Figure 1.1 Mock-Up of FRP Reinforcement System Figure 1.2 Typical cross-section in the primary direction of the proposed FRP

reinforcement Figure 2.1 A typical cross-section and assembly of a steel SIP deck form used in

conventional bridge deck construction Figure 3.1 I.F. Corporation casting bed Figure 3.2 FRP deck forms Figure 3.3 Grid with 2.5” chair and tie Figure 3.4 Longitudinal FRP bi-directional grid splice, Panel C Figure 3.5 A Beam E specimens after forms were removed Figure 3.6 Placement of concrete with chute and consolidation with vibrator Figure 3.7 Crane lifting a test specimen of the flat bed truck at SMTL Figure 3.8 Reynolds Movers move Panel B into SMTL Figure 3.9 Cross-section view of the primary direction of modified FRP reinforced

Panel C2 Figure 3.10 During the construction of Panel C2 at UW-SMTL Figure 4.1 Load contact area of Concrete Deck Panels (A, B, C, C2) Figure 4.2 Reaction frame for Concrete Deck Panels (A, B, C) Figure 4.3 A view of the concrete block supporting a concrete deck panel Figure 4.4 Series of ±0.1” LVDTs to the right of the actuator Figure 4.5 An example of a concrete strain gauge used Figure 4.6 MTS 458.20 controller, instrument power supply, Validyne signal

amplifier, multiplexer and A/C DaqBook converter Figure 4.7 Elastic test pattern for LVDT displacements of a concrete deck panel

quadrant Figure 4.8 Instrument type and location on the top surface of Panel A for the

inelastic test Figure 4.9 Instrument type and location on the bottom surface of Panel A for the

inelastic test Figure 4.10 Plan view of a positive moment beam, Beam D Figure 4.11 Positive moment beam test set-up Figure 4.12 Three-inch square bar line load assembly Figure 4.13 Instrument type and location for top surface of Beam D1 and D2 Figure 4.14 Instrument type and location for bottom surface of Beam D1 and D2 Figure 4.15 Plan view of negative moment beam (Beam E) Figure 4.16 Negative moment beam test set-up Figure 4.17 Instrument type and location of the top surface of negative moment beam Figure 4.18 Instrument type and location of the bottom surface of negative moment

beam Figure 4.19 Cross-section A-A of Figure 4.17, showing strain gauge layout

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Figure 4.20 Instrument type and location for top surface of concrete deck Panel C2 (inelastic test) Figure 4.21 Instrument type and location for bottom surface of concrete deck Panel

C2 Figure 4.22 Profile strain gauge location of section A-A of Figure 4.20 Figure 5.1 Preload concrete surface cracks at one supported end of Panel A Figure 5.2 Load versus actuator stroke for the inelastic test of Panel A Figure 5.3 Difference between the stroke and potentiometer readings, Panel A Figure 5.4 Oval concrete cracking pattern, post-inelastic test of Panel A Figure 5.5 Deflection profile of Panel A at the center span Figure 5.6 Longitudinal strains recorded along the centerline of span, Panel A Figure 5.7 Differential deflection of the middle two FRP deck forms Figure 5.8 Load versus adjusted actuator stroke for the inelastic test of Panel B Figure 5.9 View across one support showing uplift of one of the four corners and a

large vertical crack during Panel B test Figure 5.10 Difference between the stroke and potentiometer readings for Panel B Figure 5.11 Plan view of the oval concrete crack pattern on the top surface of Panel B Figure 5.12 Deflection profile of Panel B along the span centerline Figure 5.13 Material strains recorded along the centerline of span, Panel B Figure 5.14 Load versus lateral displacement opening of the center shiplap joint of

Panel B Figure 5.15 Lateral opening of the center shiplap joint for Panel B Figure 5.16 Load versus actuator stroke for the inelastic test of Panel C Figure 5.17 Difference between the stroke and potentiometer readings for Panel C Figure 5.18 View of the oval concrete pattern on the top surface of Panel C Figure 5.19 Elevation view of the bottom surface of FRP deck forms deflecting apart Figure 5.20 Deflection profile of Panel C across the center of span Figure 5.21 Material strains recorded along the center of span, Panel C Figure 5.22 Lateral opening of center shiplap joint for the inelastic test of Panel C Figure 5.23 Post-Failure Panel C Showing Punching Shear Failure Under Load Patch Figure 5.24 Load versus actuator stroke for the inelastic test of Beam D1 Figure 5.25 Depth of section versus material strain as a function of load, Beam D1 Figure 5.26 The concrete crushing near the load for both sides of Beam D1 Figure 5.27 Load versus actuator stroke for the inelastic test of Beam D2 Figure 5.28 Depth of section versus material strain as a function of load, Beam D2 Figure 5.29 Flexure-shear cracking on left and flexure cracks on the right, for Beam

D2 Figure 5.30 Two chunks of concrete dislodged at the end of Beam D2 Figure 5.31 Load versus actuator stroke for the inelastic test of Beam D3 Figure 5.32 Depth of section versus material strain as a function of load, Beam D3 Figure 5.33 Relationship of 3 concrete strain gauges at the same distance from the

centerline Figure 5.34 Material strains through the depth of section for 12” away from the

centerline Figure 5.35 Horizontal shear and flexural-shear cracking of Beam D3

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Figure 5.36 Load versus deflection curves for Span A and Span B for Beam E1 Figure 5.37 Strains from the section centered over the middle support Beam E1 Figure 5.38 Top surface strain gauge between the negative moment cracks, Beam E1 Figure 5.39 Strains through the depth of section as a function of load for Beam E1 Figure 5.40 Failure of Beam E1 for each side of beam in Span A. Figure 5.41 Load versus deflection curves for Span A and Span B for Beam E3 Figure 5.42 The first two concrete cracks over the middle support, Beam E3 Figure 5.43 Strains from the section centered over the middle support, Beam E3 Figure 5.44 Strains through the depth of section as a function of load, Beam E3 Figure 5.45 Each side of Beam E3 showing a shear failure occurring at 67 kips,

Beam E3 Figure 5.46 Concrete cracks on each side of Span A of Beam E5 Figure 5.47 Cracking at failure on side of beam without gravel epoxied to FRP form Figure 5.48 Additional cracks of Beam E5 Figure 5.49 Load versus actuator stroke for the inelastic test of Panel C2 Figure 5.50 Difference between the stroke and potentiometer readings for Panel C2 Figure 5.51 Top surface cracking pattern of Panel C2 Figure 5.52 Differential deflection of the center FRP deck forms in Panel C2 Figure 5.53 Deflection profile of Panel C2 along the center of span Figure 5.54 Strain profile of the five strain gauges, through the depth of section,

Panel C2 Figure 5.55 Other two Strain gauges 9” from center of load, along center of span,

Panel C2 Figure 5.56 Lateral displacement of center shiplap joint of Panel C2 Figure 5.57 Post mortem cut along the center of span for Panel C2 revealed the start

of a punching shear failure Figure 5.58 Preload concrete cracks for the fatigue beam Figure 5.59 Relative stiffness of the 200-kip span during the fatigue test after

various number of cycles Figure 5.60 55 kip actuator relative stiffness of 2 million-cycle fatigue test Figure 5.61 Load versus deflection curves for Span A and Span B for Fatigue Beam Figure 5.62 Cracking on each side from the inelastic test of fatigue beam Figure 6.1 Moment-curvature relationship for the positive moment beams Figure 6.2 Positive moment distribution for the elastic load of 20.8 kips for a 9’

wide panel Figure 6.3 Positive moment distribution at panel failure Figure 6.4 Load perimeter area to define bo Figure 6.5 Load versus deflection curves for all of the positive moment beams Figure 6.6 Negative moment curvature relationship centered over the middle

support, Beams E Figure 6.7 Load versus deflection curves for Span A of the negative moment beams Figure 7.1 Comparison of experimental (inelastic test) and analytical deflections at

24” from center of load along center span Figure 7.2 Deflection profile for experimental (elastic test) and analytical

measurements at center span

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Figure 7.3 Deflection profile for experimental (elastic test) and analytical measurements along A24

Figure 7.4 Moment distribution profiles over a 16’ width Figure 7.5 Negative moment profile of prototype deck for negative moment

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LIST OF TABLES

Table 3.1 The list of all 8” thick specimens constructed Table 3.2 Summary of concrete pours Table 4.1 Selected Mechanical Properties of FRP Material Table 4.2 Data record table for Panel A Ultimate Test (2/26/01) Table 4.3 Data record table for Beam D1 (4/10/01) Beam D2 (4/6/01) Table 4.4 Data record table for Beam E1 (5/9/01), Beam E3 (5/16/01), Beam E5

(5/2/01) Table 4.5 Data record table for the inelastic test of Beam E2 (fatigue beam) Table 4.6 Data record table for Panel C2 Ultimate Test (6/27/01) Table 5.1 Measured widths of the first concrete crack to develop on the top surface,

Beam E1 Table 5.2 Measured widths of the first concrete crack to develop on the top surface,

Beam E3 Table 5.3 Concrete strengths (± one standard deviation) tested according to

ASTM C-39 Table 6.1 Table of normalized curvature values to establish an approximate linear

relationship Table 6.2 Live load distribution widths for the 9’ wide concrete deck panels Table 6.3 Punching shear values and calculated capacities for the concrete

deck panels Table 6.4 Strength and effective stiffness results of the concrete deck panel tests Table 6.5 Strength and effective stiffness results of the positive moment beams Table 6.6 Strength and effective stiffness results of Span A for the negative

moment beams Table 6.7 Span-to-deflection ratios for the tested FRP specimens Table 6.8 Comparison of measured and calculated stiffness for the tested FRP

specimens Table 9.1 Design Methodologies for Steel and FRP Reinforcement for Concrete

Bridge Deck Based on 8’-8” Span Table 9.2 Performance of Positive Moment Beams to ODOT criteria for 3’ wide, 8”

thick simply supported beams

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1 INTRODUCTION

1.1 Introduction

As it can be noted from driving throughout the United States, the use of concrete

bridge decks is ubiquitous. In fact, nearly 70% of the existing bridge decks in the United

States are cast-in-place reinforced concrete (Bettigole, 1997). Many state Departments of

Transportation find the typical design life of a number of these concrete bridge decks have

been reduced, or have been exceeded (Alkhrdaji et al., 2000 and Chajes et al., 2000). Today,

as it has been for many years, the conventional method to reinforce a concrete bridge deck is

to use steel reinforcing bars. A primary cause of deterioration of a number of these concrete

bridge decks is due to the corrosion of these steel reinforcement bars (Bradberry, 2001).

Corrosion of steel reinforcement adversely affects the long-term durability of the concrete

because it leads to spalling and cracking, eventually softening the deck and leading to the

possibility of member failure (Taly, 1998).

With excessive deflections and/or the onset of member failure, these concrete bridge

decks become defined as structurally deficient. A bridge that falls into this category is either

rated for a lower load capacity, or if the damage is severe enough, taken completely out of

service (AASHTO, 2000). This has caused economic costs in rehabilitating or replacing

these bridges, as well as imposing an economic penalty to the traveling public. This also

inherently drives up the anticipated life-cycle cost. This life-cycle cost directly measures the

initial costs (like materials and installation) as well as the long-term operational costs (which

depends on such variables as expected maintenance and repair). The corrosion of steel

2

reinforcement has a negative impact upon the life-cycle costs by necessitating unplanned

costs to mitigate premature deterioration. These unintentional costs to replace prematurely

failing bridge decks (through materials, labor, relocation of traffic, etc.) and the

inconvenience to the traveling public are leading a search for a more durable material to

replace steel reinforcement.

Wisconsin Department of Transportation (WisDOT) teamed up with the University of

Wisconsin-Madison Civil and Environmental Engineering Department, Federal Highway

Administration and industry to test and implement a new technology to increase long-term

durability as well as to reduce construction time to assemble the deck reinforcement. The

new technology proposed by the UW is the use of a non-metallic glass Fiber Reinforced

Polymer (FRP) reinforcement system, to replace the use of traditional steel reinforcement.

The main advantage of using FRP is that it is a naturally inert material, and therefore,

resistant to most corrosive agents anticipated for highway application. WisDOT and

University of Wisconsin applied and were awarded funding through the Federal Highway

Administration’s Innovative Bridge Research and Construction program (IBRC) to

investigate the use of a the Stay-In-Place (SIP) FRP deck form in combination with a bi-

directional FRP top grid concrete reinforcement system.

In addition to increasing long-term durability, this FRP reinforcement system should

provide additional benefits. From an economic point of view, this FRP deck reinforcement

system is anticipated to reduce construction labor cost through ease of assembly and the

elimination of traditional cast-in-place concrete deck forms and falsework. It is anticipated

that reduced labor costs may offset the higher material cost, compared to steel reinforcing.

3

The traditional methods to design steel reinforced concrete bridge decks are not

entirely applicable with concrete bridge decks reinforced with FRP. In general, a lower

modulus of elasticity, a lack of ductility, variation in possible geometrical shapes, the

reliability of composite action, as well as other differences from design with steel, which

means that this FRP system has required experimental study. In order for the proposed FRP

reinforcement system to provide the presumed long-term durability for concrete bridge

decks, this reinforcement must satisfy the strength and serviceability requirements imposed

by the Load Factor Design (LFD) AASHTO Standard Specifications for Bridge Design, 16th

edition. Because of the relatively short time FRP materials have been on the market and

available for use in structural applications, there has not been enough time for the evolution

of a prescribed design methodology for the use of FRP in reinforced concrete bridge decks.

Therefore, in general, the purpose of this research thesis is to evaluate this proposed FRP

reinforcement system’s structural performance and determine if it can satisfy some critical

general strength and serviceability requirements for concrete bridge deck design through

laboratory testing and the use of some analytical tools.

1.2 FRP Reinforcement System and Design Methodology

The FRP reinforced concrete system is made up of two layers, a top and bottom layer

(see Figure 1.1 for picture of a FRP system mock-up). Figure 1.2 illustrates a typical cross-

section of the proposed FRP reinforcement system (cross-section cut is parallel to traffic).

The bottom tensile reinforcement is comprised of a pultruded glass FRP Stay-in-Place (SIP)

deck form spanning between, but not continuous over, the girders, which are parallel to the

traffic direction. A glass FRP bi-directional grid provides the top layer of concrete

4

reinforcement. These two glass FRP products to reinforce concrete bridge decks have been

previously studied (Bank et al, 1992 and Harik et al, 1999), but they have never been studied

together.

The FRP deck form is analogous to the main positive steel reinforcement typically

placed at the bottom of the deck perpendicular to the girder system. Each deck form is 18”

(457.2 mm) wide and overlaps with adjacent deck forms via a shiplap joint. Each of the SIP

forms are stiffened by two 3” (76.2 mm) square hollow corrugations centered 9” (228.6 mm)

apart. To ensure composite action through horizontal shear transfer between the deck form

and the concrete, ¼” (6.35 mm) aggregate is bonded to most of the horizontal surface area of

the form with Concresive 1090 epoxy. This corrugated FRP deck form has been

previously used in bridge deck construction in the Ohio Department of Transportation’s

Salem Avenue Bridge project. The performance of the FRP SIP deck form in the Salem

Avenue Bridge has been documented by Reising et al (2001). Composite Deck Systems of

Dayton, Ohio supplies the FRP deck form.

Figure 1.1 Mock-Up of FRP Reinforcement System

FRP Deck Form

Bi-Directional Grid

Plastic Chair

5

A pultruded bi-directional glass FRP grid panel (typically 4’ wide and lengths can

vary) provides the top transverse and the top longitudinal (direction with respect to the

girders) tensile reinforcement. The transverse reinforcement for the deck’s negative moment

over the girder is provided by 2” (50.8 mm) deep, “I” shaped, FRP bars. In the longitudinal

direction, the temperature and shrinkage reinforcement is supplied by oval FRP bars (dark

bars in Figure 1.1). The oval FRP bar stays in the plane of the 2” “I” bars, by penetrating

through and mechanically locking with the “I” bar web. The orthogonal members of the grid

are spaced at 4” (101.6 mm) on-center in each direction. Both bars are quite smooth and

cannot develop adequate bond to the concrete. Since they lie within the same plane and are

inherently connected, they mechanically anchor one another. The use of the pultruded grid

system to reinforce concrete slabs and bridge decks has been studied previously by Bank et

al. (1992,1993, and 1995). For the experimental phase, a smaller bi-directional grid was used

as a splice between adjacent main reinforcement bi-directional grid panels to provide

continuity of reinforcement. However, in the prototype bridge FRP reinforcement bars will

be used as a substitute to provide the reinforcement continuity. The bi-directional FRP grid

panel and splice grid are proprietary products of Strongwell Incorporated located in

Chatfield, Minnesota.

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Epoxied 1/4" GravelShiplap Joint

FRP Bi-directional Grid

FRP Bi-DirectionSplice Grid

0'-1 1/2"

8"

FRP Deck Form Figure 1.2 Typical cross-section in the primary direction of the proposed FRP reinforcement

The design methodology presented in American Concrete Institute (ACI) Guide

440.1R-01, “Guide for the Design and Construction of Concrete Reinforced with FRP Bars”

was useful to calculating reinforcing requirements (ACI, 2001). Like most ACI documents

on the design of reinforced concrete members, a strength design philosophy was adopted.

The section was designed as over-reinforced to force a concrete crushing failure, since from a

ductility point of view, this is a marginally more desirable failure than a FRP rupture.

Presumed analytical strength and service requirements were calculated assuming an effective

distribution width as provided by the LFD AASHTO Highway Bridge Design Specification,

16th edition for steel reinforced decks (AASHTO, 1996). The FRP deck form reinforcement

was assumed to be adequate for serving as tensile reinforcement to the composite section

based on previous experience. With the areas of the grid bars set for the top grid, the only

design variable left to determine was the spacing, 4” on center, of the grids for negative

moment capacity. The designed cross sectional characteristics of the proposed FRP

reinforced system is depicted in an illustration of a typical 18” wide cross section shown in

Appendix A.

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1.3 Objectives

From a structural performance point of view, the main quantitative and qualitative

objectives are as follows:

1. To examine the FRP reinforcement’s ability to laterally distribute the load in the

positive moment region and compare the experimental results to an existing

AASHTO distribution width equation that is intended for use with conventional steel

reinforced decks;

2. To determine the mode and strength at failure and evaluate the factors of safety, and

to compare the governing equation for mode of failure to the capacity measured;

3. To develop a finite element model of a FRP reinforced concrete bridge deck of the

prototype bridge and to determine negative moment distribution width for the deck

over support girders when at a load causing the concrete to start to crack over the

girder, also, to determine these loads required to initiate longitudinal concrete

cracking over the girder;

4. To investigate the effects of accelerated fatigue loading on the FRP reinforcement

system and the mechanical bond between the FRP deck form and the concrete;

5. To qualitatively determine the post-ultimate ductility of the FRP reinforcement

system;

6. To evaluate qualitative and quantitatively the difference between partial and full

coverage of the epoxy coated aggregate on the FRP SIP deck form on the composite

strength in bending; and

7. To evaluate the constructability of the FRP reinforcement components for use in

bridge construction (i.e., how well they fit together).

8

1.4 Scope

The IBRC project consists of three main phases: to investigate the strength and

serviceability behavior of the composite concrete reinforcement system though laboratory

testing, to implement the tested design into a prototype bridge deck, and finally, to complete

a long-term monitoring program. The scope of this thesis, however, is limited to

investigating the strength and some serviceability behavior through laboratory testing and

analytical modeling. The results from the testing conclude that this FRP reinforcement

system is suitable for use in the prototype bridge and for bridges in general. Therefore, it

will be implemented in the construction of one part of a new twin bridge structure owned by

the Wisconsin Department of Transportation. The other twin of the new bridge structure, on

USH 151 near Waupun, Wisconsin, will use conventional steel reinforcement for the bridge

deck, thus providing a hands-on opportunity to compare performance and durability of the

two materials in service. A detailed description of the prototype bridge structure is provided

in Appendix A.

The scope needed to achieve the objectives of this project is outlined as follows:

1. Constructed and experimentally tested, through 3-point bending with a load footprint

of approximately equal to the contact area of a double wheel tire, four FRP reinforced

concrete deck panels to determine the distribution width, mode of failure and strength

factors of safety.

2. Constructed and experimentally tested, through 3-point bending, three FRP reinforced

concrete beams to isolate and determine the positive moment bending characteristics

of the concrete/FRP deck form composite section.

9

3. Constructed and experimentally tested three 2-span continuous FRP reinforce

concrete beams to isolate and determine the negative moment bending characteristics

of the concrete/FRP bi-directional grid panel composite section.

4. Constructed and experimental tested one 2-span continuous FRP reinforce concrete

beam to isolate and determine the fatigue resistance of the FRP materials and the

mechanical bond between the FRP deck form and the concrete.

5. Used the experimental values from the above tests to verify and construct a finite

element model for a section of the prototype bridge that better replicated the scale,

continuity, and end conditions expected in the actual prototype bridge deck. This was

done to determine the negative moment distribution width and the load that produced

concrete cracking over the girder.

Not apart of this scope is the formal comparison of service load deflections of the

FRP reinforced section to AASHTO serviceability criteria. Harik et al (1999) studied the

AASHTO LRFD service limit state for the service load deflection performance of a concrete

section reinforced with the same SIP FRP deck form for the tensile reinforcement, in

combination with, Glass Fiber Reinforced Polymer (GFRP) rebar for the compression

reinforcement.

10

2 LITERATURE REVIEW

2.1 Introduction

Fiber Reinforced Polymer high-performance materials are being diversely used for

structural applications in highway bridges. This chapter presents the literature review of

some previous studies for the use of FRP materials, specifically for highway bridge deck

applications. Most papers reviewed in this chapter address, in some part, the design

performance of their specimens according to AASHTO bridge design specifications. First

part of the chapter will review a few studies where the entire superstructure system is

composed of an FRP modular core slab supported on steel girders, and even in one case, a

study where a FRP core system alone acts as the highway superstructure. The next portion of

the literature review discusses selected studies where FRP bars are implemented as the

concrete composite reinforcement for slabs. Another area of the literature review will

examine studies of the same, or similar, concrete FRP reinforcement grid studied in this

thesis. Also included in this literature review is a look into the conventional use of non-

composite steel SIP forms with reinforced concrete as a bridge deck. The final part of the

literature review examines a couple of studies where the same FRP SIP deck form studied in

thesis has been employed as the tensile reinforcement for composite concrete slabs.

2.2 Bridge Decks Constructed from FRP Composites

Recently, a number of states have begun constructing bridges where the entire

structural deck system is comprised of FRP composites. A fully FRP bridge deck has many

beneficial traits over a conventional steel reinforced concrete deck, such as being lighter,

11

corrosion resistant and a higher strength-to-weight ratio. Delaware Department of

Transportation (DelDOT) has replaced an existing concrete slab, on steel girders, with a

lightweight glass FRP deck. Chajes et al (2001) detailed the rehabilitation of this existing,

low-volume, 35 ft clear span bridge and then documented the performance of a field load

test. The design requirements called for a 10-inch thick GFRP slab, which weighs 30 lb/ft2,

to continuously span over six longitudinal girders spaced at 33.8 inches. For a wearing

surface, the slab had a shop applied polymer concrete finish. Although the paper did not

address the design details, it did hint that the polymer composite slab was governed by

deflection limits.

The field test was conducted with a fully loaded 10-wheel dump truck and both the

performance of the GFRP deck and the girders were of interest. Three different load

applications were completed: static, slow roll and dynamic. The test results for

displacements, peak strains, bending of the girders, transverse load distribution, longitudinal

and transverse bending of the deck were recorded. According to the test results, the GFRP

composite bridge slab performed well within the serviceability and strength requirements.

Of specific interest, the superstructure achieved a deflection ration of L/2100 (limit

recommended by code is L/800) without composite action between the deck and the girders.

One design aspect the Chajes et al (2001) study did not address, because of low-

volume traffic, was the long-term fatigue performance of the FRP composite slab. In order

to jump from low to high volume traffic use, the fatigue performance of these FRP composite

slabs needed to be experimentally documented. A study by Lopez-Anido (1998) investigated

the long-term performance of a modular FRP composite bridge deck, similar to the Chajes et

al system, under cyclic loading. The Lopez-Anido study examined modular FRP deck panels

12

supported on longitudinal girders. The FRP composite deck system performed well, without

major degradation in stiffness or strength, when it was subjected to 2 million load cycles of

an HS20-44 design truck adopted from the LRFD design philosophy for highway steel

structures. These results help provide some confidence for FRP as a competent material to

provide fatigue resistance.

A step beyond the use of FRP composites to act as the deck supported on a steel or

concrete stringer superstructure system is to employ FRP composites as the superstructure

itself. The New York State Department of Transportation replaced a significantly

deteriorated 25-foot span, steel reinforced concrete slab bridge (ADT of 300 vehicles per

day) with a two-piece superstructure cell core system that provides stiffness in two

directions. The two-foot thick cell core system was designed according to AASHTO strength

and serviceability criteria. After the original superstructure was removed, the two-piece

system was installed within six hours and the bridge was reopened to traffic a few weeks

later. Before being opened to traffic, the bridge was proof tested with HS25 truck loadings to

check the structure’s integrity, to establish base line condition and to compare actual

performance with theoretical calculations. Proof testing concluded that the load capacity was

greater than what was determined by analytical analysis and the bridge’s span to deflection

ratio was L/2010. This study also determined that FRP composites could be a cost-effective

alternative for short span bridge superstructures.

2.3 Concrete Bridge Decks Reinforced with FRP Bars

Probably the most natural use of FRP materials to reinforce concrete bridge decks is

to replace conventional steel reinforcing bars with FRP bars. Bradberry (2001) of the Texas

13

Department of Transportation (TexDOT) documented the structural design of a concrete

bridge deck with the top mat of concrete reinforcement made of GFRP bars. The designed

nominal deck thickness was 8 inches and was supported by five prestressed concrete girders

spaced at 7.84 feet, center-to-center. The main design issues discussed, relative to this thesis,

were design ultimate strength, calculated crack width and shear strength of the concrete

bridge deck designed with GFRP bars. Two main conclusions from this paper were: the

author found that the bar spacing in the top GFRP mat was controlled by the serviceability

issue of crack width and in the author’s judgment, the significant lack of ductility displayed

by FRP bars was not a barrier for their use in concrete bridge decks. The conclusions exhibit

the main weaknesses for use of FRP bars, the lack of stiffness and ductility. Although, the

FRP bars do provide a way to prevent deterioration of concrete bridge decks where steel has

performed inadequately.

Bradberry (2001) did state concerns about unanswered question about long-term

durability issues with GFRP. Kumar et al (1998) did conduct a study on GFRP reinforced

concrete deck-stringer systems under fatigue loading. They measured the rates of

degradation of strength and stiffness and compare the performance of GFRP concrete decks

to previous research that used steel reinforcement. The main conclusion is that FRP

reinforcement bars performed well, when compared to steel, in the crack propagation zone.

This study also found that 2 million fatigue cycles, with a load at about 20% of ultimate

strength, could be conservatively assumed to be 80% of the design life of the FRP bars.

These results provide some confidence that FRP bars can safely satisfy fatigue serviceability

requirements for concrete reinforcement of bridge decks.

14

2.4 Concrete Bridge Decks and Concrete Slabs Reinforced with FRP Grids

Survey of the literature addressing FRP composites to reinforce concrete bridge decks

reveal that the main alternative to FRP bars is the use of FRP grids or gratings. Bank et al

(1992, 1993, and 1995) conducted some of the earliest testing that investigated the

performance of FRP gratings. The Bank et al study (1992) investigated full size bride deck

panels reinforced with two different commercially provided orthogonal gratings (Safe-T-

Grate and Duradek) along with control slabs reinforced with steel bars. Each full size

slab was designed according to AASHTO strength and serviceability recommendations using

a nominal HS-25 wheel loading. The slabs were only reinforced in the bottom layer to test

positive moment behavior and failure. As designed, the FRP slabs failed due to concrete

compression failure and the controls specimens failed due to the yielding of steel bars (and

then concrete crushing). This study helped demonstrate the behavior of the different failure

modes between the FRP and steel reinforcement. Experimental service load deflections of

the FRP reinforced slabs were close to the analytical values calculated. Also, the safety

factors of ultimate strength to service load were greater than three. In addition, the ultimate

loads of the FRP slabs were greater than the load that caused yielding in the steel reinforced

control slab. This study provided early evidence that FRP gratings could be a workable

reinforcement alternative for bridge decks.

In 1993, Bank and Xi investigated analytical and experimental agreement of concrete

slabs reinforced with FRP gratings using the same gratings from the 1992 study. Dependent

upon the modulus of elasticity of the FRP, they concluded that the analytical models

illustrated acceptable predictions. The accuracy of the slab stiffness, however, is dependent

15

upon the accuracy of the value for the FRP modulus of elasticity. Another conclusion, at that

time, is that addition study is required to predict the shear strength of these slabs.

Bank and Xi, in 1995, evolved their research to investigate the shear, or punching

shear, capacity of the concrete slabs reinforced with FRP gratings. Punching shear capacity

of a reinforced concrete slab-stringer bridge system is important because the generally

accepted failure of a concrete deck is due to punching shear (exception to overhangs) (Bank

and Xi, 1995 and Bradberry, 2001). In this study, punching shear capacity was tested with

variables of: grating bar spacing, grating panel layout, and panel splice detail. Although

there were no steel reinforced slabs tested for comparison, the main conclusion is that the

punching failure behavior was similar to what is typically seen for a steel reinforced slab. In

addition, since concrete alone provides the punching shear resistance, the variables play little

role to capacity, although they did influence the post-ultimate behavior.

A more recent study by Matthys and Taerwe (2000) examined flexural and punching

shear behavior of concrete slabs reinforced with two types of NEFMAC grids (carbon and a

mixture of carbon and glass fibers). With the FRP slabs designed to be comparable to a

reference steel reinforced slab, this study provided a direct comparison of performance

between FRP and steel. From a bending perspective, the authors conclude that the benefit of

the added strength of the FRP grids is only partially used since the controlling design criteria

was for serviceability deflections. They also conclude that the safety factors for strength are

higher, however, the ductility is lower than steel. From a punching shear capacity

perspective, two main conclusions stand out. First, the FRP slabs, with comparable

reinforcement strength as the steel, showed lower punching shear resistance and stiffness

(although with an increased reinforcement ratio and or depth of section this can be

16

overcome). Second, introducing an equivalent reinforcement ratio (to take into account the

axial rigidity of the reinforcement), ρE/Es, into the analytical predictions from a variety of

international code equations provided satisfactory factors of safety for the experimental

results of the punching shear capacities.

2.5 Use of Steel SIP Forms with Reinforced Concrete Bridge Decks

The conventional use of steel stay-in-place deck forms for bridge applications is

commonplace. They are used, however, only for support of wet concrete between girder

spans and are not applicable to replace or reduce the bottom transverse steel reinforcement.

They are not intended nor are they permitted to act compositely with the concrete to provide

strength after the concrete has set. The main reason they are not permitted to be used as

reinforcement is because they are subject to long-term corrosion problems. Their main

purpose is to facilitate construction by reducing labor through the elimination falsework and

formwork.

The steel stay-in-place forms can be used for steel or concrete girders. Figure 2.1

shows a typical cross-section and assembly of a steel SIP form for a steel stringer or girder

bridge deck system. They are not composite for bridge applications, however, they can be

used as composite reinforcement in building applications. In concrete floor design a steel

SIP deck form alone can provide enough reinforcement in positive moment regions,

however, in building applications these forms are generally isolated from exposure to

moisture.

17

A few of the larger manufacturers of steel SIP deck forms for building and bridge

applications are United Steel Deck, Inc. located in South Plainfield, NJ and Epic Metals

Corporation from Rankin, PA.

Depth

Pitch(varies)

Coverage(varies)

Top Flange(varies)

Stringeror Grider

Stringeror Grider

Deck

Supportangle leg

1"min. bearing

1/2" flange widthSpan 1"

A

A

Section A-A

Figure 2.1 A typical cross-section and assembly of a steel SIP deck form used in conventional bridge deck construction

2.6 Concrete Bridge Decks and Concrete Deck Panels Reinforced with FRP SIP Forms

The use of the SIP FRP deck form to provide a form for wet concrete as well as the

tensile reinforcement for a composite concrete deck is a relatively new concept. Harik et al

(1999) have pioneered the experimental investigation of a similar GFRP deck form (18” wide

each, stiffened with two corrugations) that was studied in this thesis. The Harik et al study

investigated three-point bending of the SIP FRP form/concrete composite deck panels, with

FRP bars acting as compression reinforcement, designed according to ASSHTO LFRD

18

(1998) and LFD (1996) specifications. Their goal was to load three composite concrete deck

panels, of varying depths and spans, with an AASHTO standard HS25 truck wheel load and

observe the service load deflections, the performance to cyclic loading, the strength capacity

and the mode of failure.

The failure mechanism observed from these three-point bending tests was the same as

the one observed in the testing conducted for this thesis. The positive moment test revealed a

flexure-shear cracking and a de-bonding of the FRP deck form from the concrete at failure.

The service load deflections of all specimens were well below the allowable. In addition, the

panel stiffness did not diminish after being subjected to cyclic loading. Finally, the factors of

safety were greater than three for all specimens. Therefore, as designed according to

AASHTO criteria, the selected strength and serviceability requirements were satisfied.

These results breed confidence that the SIP FRP deck form can easily satisfy the

serviceability deflection criteria.

The Ohio Department of Transportation was confident enough with the FRP concrete

reinforcement system used in the Harik et al study that they implemented it into an actual

bridge and Reising et al (2001) documented its performance. The ODOT Salem Avenue

Bridge Project retrofitted a 5-span bridge with four different FRP composite panel systems

and one of the four systems was the SIP FRP deck form/bar concrete reinforcement. From a

global performance perspective, Reising et al found that this system did not reduce the

overall stiffness of the bridge. Additionally, such values for girder distribution and impact

factor were within the recommendations according to AASHTO specifications.

19

3 PREPERATION OF TEST SPECIMENS

3.1 Introduction

This IBRC project required the construction of concrete bridge deck panels and

beams reinforced with FRP for a series of serviceability and strength testing. I.F.

Corporation, on east side of Madison, Wisconsin, was contracted by the UW to prepare,

assemble, pour and deliver pre-cast deck test specimens. This chapter documents the test

specimens, the casting facility, assembly methodology, concrete specifications, transportation

and an overall evaluation of the deck test specimen preparation at I.F. Corporation. Also

documented, is the construction of an additional concrete deck panel at the Structures and

Materials Testing Laboratory (SMTL) on the University of Wisconsin–Madison Engineering

Campus.

3.2 Test Specimens

It was decided that 11 deck specimens were required to determine the stated

objectives in the Introduction. Table 3.1 details the specimens that were constructed for

laboratory testing. The laboratory testing did not require the inclusion of a wearing surface;

therefore, UW Professor Michael G. Oliva designed all specimens with 1.5” of concrete

cover above the bi-directional FRP grid panel. The chapters to follow will address in more

detail the purpose of the test specimens. Appendix B illustrates the location of strain gauges

applied during the assembly process, for some of the test specimens. Panels A, B and C were

fitted with 6 strain gauges each, located on the bottom of the FRP deck forms, for one-half of

the symmetric concrete deck specimen (2 per panel) along the intended center-line of the

20

span. Three of the negative moment deck specimens, Beams E, were fitted with one strain

gauge each, centered on the web of the “I” beam of the bi-directional grid, centered over the

intended support (See Appendix B for gauge locations).

Table 3.1 The list of all 8” thick specimens constructed

Specimen Category Name Span Purpose Panel A 11’-6” Panel B 9’-10” Panel C 8’-0”

Concrete Deck Panels (all 9’-0” wide)

Panel C2 8’-0”

To determine distribution widths, mode of failure and strength

safety factors

Beam D1 Beam D2

Positive Moment Beams

Beam D3

All 8’-8” Test the positive moment properties of the corrugated deck form

Beam E1 Beam E3 Beam E4

Two Span Negative Moment Beams

Beam E5

Each span 8’-0”

Test the negative moment properties of the bi-directional grid over a support

Two Span Fatigue Beam Beam E2 Each span 8’-0”

Test the fatigue strength of the FRP system and the mechanical bond

Figure 3.1 I.F. Corporation casting bed

21

3.3 Precast Facility

At I.F. Corporation, a 9’ x 80’ concrete pre-cast forming bed was utilized to prepare

the deck specimens. The casting bed seen in Figure 3.1 had the ability to form concrete

panels up to 8” thick. The pre-casting bed geometry would not allow a single pour for all

specimens; therefore two separate pours were completed. For curing, the casting bed was

enclosed and heated at 60 degrees for up to 3 days after casting.

Figure 3.2 FRP deck forms

3.4 Assembly Methodology

After assembling the FRP reinforcement for the first concrete deck panel, Panel A, it

was quickly learned how to expedite the process. A sequential order was developed through

experience. First, the pour bed was swept clear of all debris to ensure the FRP deck forms

could lay flat. Next, as seen in Figure 3.2, duct taped was used to enclose the FRP deck form

corrugation end openings to prevent concrete from migrating inside the voids during concrete

22

placement. At this point, lap joints between the adjacent forms were connected together with

screws at approximately 5’ intervals. This was a safeguard procedure to ensure stability of

the forms for the concrete pour only and was not intended to provide any structural

continuity. Preparing ahead for tie down support of the FRP bi-directional grids, wire ties

were anchored to the top face of the FRP deck form corrugations with screws at

approximately 4’ intervals. Next, the FRP grids were placed, along with transverse splices,

into position and propped up with 2.5” plastic chairs. As seen in Figure 3.3, the plastic chairs

were placed upon the top surface of the corrugations.

Figure 3.3 Grid with 2.5” chair and tie

From experience it was found that the FRP grids were sufficiently stiff to walk upon

with chair spacing intervals of 3’ to 4’. Once the grids were into position, they were tied

down with the previously anchored wire ties. Then the transverse splice grids seen in Figure

3.4, used to supply temperature and shrinkage reinforcement continuity between the main bi-

23

directional grids pieces, were tied tightly below the main grids. The assembly process

provided a stable system to withstand walking on and for concrete placement.

Three beams, D1 thru D3, of the same length were assembled to test the positive

moment properties. Beams D1 and D2 had top FRP bi-directional grids, while D3 was cast

without this FRP grid to compare results. These beams did not contain any grid splices;

therefore they were constructed quite simply with one FRP grid and two, side-by-side deck

forms. The order of assembly of the FRP reinforcement components mimic that which was

derived for Panels A, B and C.

Figure 3.4 Longitudinal FRP bi-directional grid splice, Panel C

The negative moment properties of the FRP reinforced system were tested with four

E-type deck beams (see Appendix B for geometry and layout). Three of the four negative

24

moment beams had gauges (E1, E3 and E5) to measure top layer grid strains of the main

reinforcement over the intended support. The remaining negative moment beam, E4, had no

bi-directional grid reinforcement over the intended support, to provide results for negative

moment capacity of an un-reinforced section. The remaining beam, E2, had the FRP grid

reinforcement over the support, but was intended for fatigue testing.

The Beam E specimens were spliced longitudinally with FRP 8” splices at each

longitudinal, bar made-to-fit over the top flange. Theses splices did not provide any strength

capacity, but rather they were to provide proper alignment between grids. A special aspect of

the positive reinforcement scheme for Beams E was an 8” discontinuity or gap of the FRP

deck forms over the support, since the forms simply span between girders and aren’t

continuous over the girders. Photo 5 shows a Beam E right after the forms were removed.

Figure 3.5 A Beam E specimen after forms were removed

25

3.5 Concrete Placement and Specifications

The concrete mixed used was a Wisconsin DOT Class D mix (see Appendix B for

mix design) typically used by the WisDOT for bridge decks (WisDOT, 1996). The concrete

design strength was dropped, however, from the typical 4500 psi to 4000 psi. A 7” slump

was specified to increase the concrete workability and to minimize honeycomb pockets. It

was noticed that one drawback of the FRP grids was that the main reinforcement “I” beams

could easily cause void pockets without proper consolidation. These pockets typically

formed between the outside “I” beam and the concrete forms that run parallel to the main

reinforcement. An additional measure taken to help prevent this problem was to hold the

maximum aggregate size to ¾”.

Figure 3.6 Placement of concrete with chute and consolidation with vibrator

The ready-mix concrete was supplied by M and M Concrete and was delivered from

the truck by chute and spread with a shovel (Figure 3.6). During the placement, a concrete

pencil vibrator facilitated the concrete movement between tight places and ensured good

26

consolidation. After striking off extra concrete, a float finish was applied to the surface of

each specimen (The smoother the finish, the easier it is to see cracks!). Three cylinders were

cast for each test specimen with an additional three for each pour, of which there were two

separate pours.

On January 12th, 2001, Panels A, B and C were poured. About two weeks later, on

January 22nd, the outstanding beams were poured. The original plan was to have 48 hours at

60 degrees for curing. After testing a 2-day cylinder from the first pour, however, the

compressive strength was around 1640 psi, which was below the 2000-psi strength required

before the beams were transported to SMTL. Therefore, for the second and last pour, the

cure time was increased to 60 degrees for three days. After three days, a cylinder tested at

3380 psi, a significant increase for an additional day of curing. Even though the design

mixes are the same, they were from different batches of concrete, which may also have had a

slight impact on the additional strength.

Figure 3.7 Crane lifting a test specimen of the flat bed truck at SMTL

27

3.6 Transportation

I.F. Corporation was also contracted to load and deliver the test specimens, via a flat

bed semi tractor-trailer, to SMTL. As seen in Figures 3.7 and 3.8, a crane provided by

Reynolds Movers, whom also moved the test specimens into SMTL, unloaded the specimens.

Figure 3.8 Reynolds Movers move Panel B into SMTL

3.7 Evaluation

The casting process at I.F. Corporation and the delivery of the test specimens was a

success. Beam E4 (negative moment beam without top FRP grid reinforcement) cracked in

half at the centerline, transversely, on the trip to SMTL. This test specimen had absolutely

no reinforcement within the center 8” of the beam. Since the beam lifted onto the truck at

I.F. Corporation without difficulty, the beam most likely cracked due to the vibration caused

by the ride on the back of the truck.

28

3.8 Construction of an Additional Concrete Deck Panel

After the first phase of testing, the original FRP reinforcement system displayed two

behavioral concerns, partial composite action and a susceptibility to longitudinal concrete

cracking above the shiplap joint. Therefore it was decided, based on these initial test

observations, that an additional concrete panel should be constructed and tested with a

modified FRP reinforcement system to address these concerns. The two modifications to the

FRP reinforcement system were: additional aggregate coverage to all horizontal surfaces of

the FRP deck panel to improve the composite action and the addition of a 4” wide FRP

Fibergrate Molded Grating, 2” square pattern, to reinforce the concrete directly above the

FRP deck panel shiplap joint. Figure 3.9 depicts a typical cross-sectional view of the

modified FRP reinforced section of concrete Panel C2 (transverse to the corrugations). This

additional concrete panel was constructed at SMTL.

Shiplap Joint


FRP Bi-DirectionSlice Grid 2" Square

FRP Grating

Epoxied Gravel

FRP Deck Form Figure 3.9 Cross-section view of the primary direction of modified FRP reinforced Panel C2

This methodology of assembly of the reinforcement parts was very similar to the

process outlined earlier in this chapter. The only difference was the panel was formed with

plywood buttressed with 2” x 6” whalers. Figure 3.10 shows Panel C2 during the

29

construction process; it also shows the additional coverage of epoxy-aggregate on the FRP

deck forms and the inclusion of 4” wide x 1” deep FRP Fibergrate Molded Grating placed

above the corrugation valleys with the shiplap joints.

Panel C2 was equipped with internal strain gauges applied directly to the FRP

components before the concrete was poured. These were included to allow the measurement

of strains through the depth of the reinforced cross section. The exact same concrete mix

used for the original test specimens was ordered from the same company, M&M Concrete.

Figure 3.10 During the construction of Panel C2 at UW- SMTL

3.9 Concrete Quality Control

For concrete quality control, three 28-day concrete test cylinders were poured for

each specimen cast. In addition, from each batch of concrete (i.e., each concrete truck) an

additional three concrete cylinders were made. Totally, there were three concrete batches:

30

one for Panels A, B and C; another for all Beam D, Beam E and Fatigue Beam specimens,

and one more for Panel C2. Table 3.2 summarizes the specifics of each of the concrete

pours. The mix design for the WisDOT Class D concrete is shown in Appendix B.

Table 3.2 Summary of concrete pours

Specimens Date Poured

Mix design #208, M&M Concrete, WDOT Class D

Cylinders Made

Panels A, B, C 01/12/01 4000 psi/7” Slump/ ¾” Agg. 12 Beam D, E1 and

Fatigue 01/22/01 4000 psi/7” Slump/ ¾” Agg 18

Beam E3,E4, E5 01/22/01 4000 psi/7” Slump/ ¾” Agg 12 Panel C2 05/23/01 4000 psi/7” Slump/ ¾” Agg 3

31

4 TEST METHODS

4.1 Introduction

To characterize the FRP reinforcement system, two primary areas of experimental

research were conducted. The first area, the focus of this thesis, was the structural testing of

the composite FRP reinforced deck panels and beams. Although not covered in the scope of

this thesis, a second (concurrent to the structural testing) investigation into the physical and

mechanical characterization of the FRP material was conducted by Dietsche (2002). The

results of selected mechanical properties of the SIP deck form and bi-directional grid are

shown in Table 4.1. This second investigation provided mechanical properties used to

calculate analytical deflections and strengths of the structural specimens.

Table 4.1 Selected Mechanical Properties of FRP Material

Deck Material Test Configuration Strength MPa (ksi) Modulus GPa (ksi) Longitudinal Tension 504.3 ± 111.6 (73.2 ± 16.2) 33.4 ± 4.1 (4841 ± 593.6) Longitudinal Compression 484.4 ± 91.6 (70.3 ± 13.2) 37.5 ± 4.9 (5436 ± 706.6) Transverse Tension 29.6 ± 1.30 (4.3 ± 0.19) Not measured Transverse Compression 82.0 ± 6.5 (11.9 ± 0.94) 7.5 ± 0.58 (1087 ± 84.0) Short Beam Shear 38.6 ± 3.3 (5.6 ± 0.48) Not Applicable Grid Material Test Configuration Strength MPa (ksi) Modulus GPa (ksi) Longitudinal Tension 571.9 ± 12.3 (83.0 ± 1.79) 30.9 ± 1.07 (4480 ± 155.7) Longitudinal Compression 648.3 ± 5.85 (94.1 ± 0.85) Not measured Transverse Tension 73.7 ± 7.51 (10.7 ± 1.09) Not measured Transverse Compression 170.9 ± 3.24 (24.8 ± 0.47) Not measured Short Beam Shear 51.0 ± 0.90 (7.4 ± 0.13) Not Applicable

Note: Values included ± 1 standard deviation. Test data based on a minimum of 5 specimens.

The laboratory testing was carried out at the SMTL. This structural testing was

broken down into a series of three different tests and a fatigue test. The sequence of testing

was as follows: 3 concrete deck panels (Panel A, B, C), 3 positive moment beams (Beams

32

D), 3 negative moment beams (Beams E), 1 additional concrete deck panel (Panel C2) and

finally a fatigue beam (Beam E2).

4.2 Concrete Deck Panel Test

In order to establish an understanding of the general structural behavior of an actual

FRP reinforced concrete bridge deck, FRP reinforced concrete deck panels were constructed

(described in Chapter 3) and tested. The concrete deck panels were each 9’-0” wide and 8”

thick, but the span length of each panel was different. Simply supported spans of 11’-6”, 9’-

10” and 8’-0” (Panels A, B, C, respectively) were selected to allow a regression analysis of

various structural behaviors dependent on span length. Each concrete deck panel was

subjected to a load contact area shown in Figure 4.1, 8” x 23”, approximately equivalent to

the contact area of a double tire wheel load of 18.4 kip (AASHTO, 1996). AASHTO defines

the load contact area as a function of load; however, this area cannot easily be adjusted

during testing. The purpose of testing these concrete deck panels (Panel A, B, C) was to

acquire information on the elastic behavior, failure mechanism, load distribution width and

ultimate capacity.

33

9' Width

CL

8"

23"

Concrete Panel Span(Varies)

Contact Area of Load to Simulate Dual Wheels

Direction of TrafficG

irde

r Loc

atio

n

Gir

der L

ocat

ion

Figure 4.1 Load contact area of Concrete Deck Panels (A, B, C, C2)

4.2.1 Concrete Deck Panel Test Station

A reaction frame was assembled to mount the actuator system needed to vertically

load the concrete deck panels. The reaction frame was anchored and prestressed onto

concrete blocks that attached to a structural floor. The columns and beam of the frame were

made from stiff steel sections. The entire frame system, seen in Figure 4.2, was prestressed

into a structural floor system in order to prevent any vertical displacement under the

anticipated loads. Any vertical displacements during testing would have affected the

accuracy of the actuator stroke (displacement) and actuator control.

34

Figure 4.2 Reaction frame for Concrete Deck Panels (A, B, C)

Supporting the concrete deck panels were two concrete blocks, 20” high by 20” thick

by 108” long, one at each end, as seen Figure 4.3. A layer of plaster was sandwiched

between the concrete block support and the structural floor to ensure an even distribution of

reaction on the floor and to prevent any rocking of the support during testing. A 5-ton

overhead crane was used to position the concrete deck panels on the concrete support blocks.

The concrete deck panels were placed upon an 8” wide x ¼” to ½” thick layer of plaster.

Like the plaster between the support blocks and the floor these plaster bearings were cast to

evenly distribute support loads to the deck. The bearing also provided clearance for rotation

of the concrete deck panel during testing.

35

Figure 4.3 A view of the concrete block supporting a concrete deck panel

A 200 kip capacity MTS closed-loop, servo-hydraulic actuator provided loading to the

concrete deck panels. The 200 kip hydraulic actuator can be seen in Figure 4.2, mounted in

the center of the beam of the loading frame. An MTS 458.20 digital servo control station

directly controlled the 200 kip hydraulic actuator. The load was applied from the actuator

base plate (13” square) onto a 2” thick steel plate (8” x 23”) then through a 2” thick

elastomeric rubber pad (12” x 24”) to the top surface of the concrete deck specimen.

4.2.2 Data Acquisition and Instrumentation

To obtain reliable and accurate data from experimental testing, calibrated

instrumentation and a data acquisition system was essential. The paragraphs to follow are a

general description of how the instrumental data from the experiments was collected,

converted to useful form and recorded. From the previously mentioned MTS 458.20 digital

36

controller (which can be categorized as instrumentation) came two elementary channels of

data output, the load and the stroke. This MTS control unit sent voltage signals to the

actuator for control and directly to the data collection multiplexer for recording applied load

measured by the actuator load cell and the stroke.

To measure specimen deflections at desired locations, Schaevitz AC 100 and AC

1000 Linear Variable Displacement Transformers (LVDT) and a Genisco Technical

potentiometer were used. An LVDT, see figure 4.4, consists of an electromagnetic coil in the

cylinder and a plunger. The cylinder was isolated from the specimen by mounting it on a

stationary reference frame. The plunger is considered the active piece of the instrument. It

was inserted into the barrel of the cylinder with one end resting at a selected position on the

surface of the specimen. As the specimen deflected, the plunger moved with it and thus

changed its position in the barrel of the stationary cylinder. This change in position of the

plunger relative to the stationary cylinder was measured by varying coil voltage and a signal

as voltage was sent from the cylinder to the data acquisition system and recorded.

37

LVDT’s

Figure 4.4 Series of ±0.1” LVDTs to the right of the actuator

To ensure the LVDTs provided reliable and accurate data they were calibrated before

use. The calibration process provided a calibration factor that was used to convert the

voltage sent by the LVDT to an actual deflection value recorded by the data acquisition

system. Each LVDT had a range of accurate deflection measurement. For this project, two

LVDT deflection ranges were required, ± 1.0” (AC 1000) and ± 0.1” (AC 100). Regardless

of the LVDT range of measurement used, the general calibration process can be described as

follows: the cylinder was mounted and kept stationary. Using a voltmeter, the plunger was

inserted and positioned to one end of the range (for example, either +1” or –1”). Then

precision instruments, either certified gage blocks or a micrometer, were used to move the

plunger through a known distance and a change in voltage was noted. A series of known

distances were measured through the range of the LVDT and a linear relationship was

38

established between changes in distance versus a change in voltage. Thus, a calibration

factor was derived in units of distance per volt, or inches per volt.

A Genisco potentiometer, model PT-101A, with a 4” range was used to determine

deflections near the center of the concrete panel test. The potentiometer measures

displacement differently than an LVDT, but it has a very similar calibration procedure as

described for an LVDT. The calibration factor determined by the calibration procedure was

in units of distance per volt, or inches per volt.

In addition to deflection data, strain data were recorded during testing. For all tests

conducted, Micro-Measurements Precision Strain Gauge type CEA-06-250UN-120 was used

for all FRP surfaces and type EA-06-20CBW-120 was used to measure strains for all

concrete surfaces. Strain gauges were first cemented at a desired point on the surface of the

specimen. Figure 4.5 shows an example of a strain gauge on the concrete surface of one of

the concrete deck panel specimens. In general, strains were determined by measuring a

change in voltage due to change in resistance from the strain gauge during the test. As the

strain gauge stretched or contracted, the voltage changed. This voltage was converted to a

strain by a calibration factor. Each strain gauge had a known calibration factor of strain per

volt. However, like the LVDT, strain gauges must be calibrated to derive the calibration

factor. The general procedure to calibrate the strain gauges used can be described as follows:

the output voltage signal from the strain gauge was zeroed. Next, a resister of a known

resistance was used to shunt the wires attached to the strain gauge. At this point, the output

voltage through the wires to the strain gauge and the resister were read. This value was used

to derive the known adjustment factor of the strain gauge. This result is known as the

39

calibration factor, strain per volt, and was used to convert the voltage output from the strain

gauge to strain data (inch/inch).

Figure 4.5 An example of a concrete strain gauge used

The LVDTs, potentiometers and the strain gauges required a power supply or voltage

input. In addition, the values of the voltage output (data) from these instruments were quite

low and necessitated amplification. A Validyne instrument power supply and signal

amplification unit carried out these two tasks. This was a nerve center, data feedback

collector and organization point of the instrument network (see Figure 4.6). Each instrument,

whether it was a LVDT, potentiometer or a strain gauge, required a set of wires to be

connected to a designated channel within the Validyne unit.

40

Figure 4.6 MTS 458.20 controller, instrument power supply, Validyne signal amplifier,

multiplexer and A/C DaqBook converter

From the Validyne unit came the amplified voltage signal. This amplified voltage was

routed to a multiplexer unit to properly channel these voltage signals through a DaqBook

component. The DaqBook unit took these voltage signals and converted them from an

analog to a digital signal. A digital voltage signal was required for the computer software

program. LabView computer software, installed on an IBM compatible laptop computer,

was used to read the digital voltage output and converted these values to the desired output

data, such as displacement or strain. This conversion was where the calibration factors

played their roles. An additional task for the LabView software was to record the data to a

file.

41

4.2.3 Instrumentation Applied and Data Recording

The procedure to test each concrete deck panel involved three sequential steps:

conditioning, elastic testing, and finally the ultimate or inelastic testing. The paragraphs to

follow are the description of the purpose, kind of and location of instrumentation and how the

data was recorded for each of these steps.

To replicate the actual in-service state of a concrete deck reinforced with FRP (i.e.

cracked section), each deck panel was first conditioned. Conditioning involved loading the

specimen to a level above the expected service load. In this case, each deck panel was

subjected to ten cyclic loadings from 0 to 24 kips. The loading could be classified as

“somewhat” sinusoidal because a strict programming loading sequence did not carry out the

load application, but rather, the pace was controlled by hand. During this step,

instrumentation was used and data was recoded, but only as a system check and as a backup

in case of premature failure occurred.

After the conditioning step, the next experimental step conducted on each concrete

deck panel was an elastic test. This elastic test involved measuring the deflections on the top

surface of the concrete deck panel under service wheel load, 16 kip, in a quarter section of

the deck surface. Because of natural symmetry of the concrete deck panel, measuring

deflections over a quadrant provided a satisfactory deflection profile for the entire surface.

Figure 4.7 depicts the quadrant grid pattern used to determine deflection for the entire surface

of the concrete deck specimen. To measure the relatively small deflection under a 16 kip

load, LVDTs with a ± 0.1” range were used. The LVDTs were mounted on a reference

frame to isolate them from movement of the specimen. The reference frame itself was pin

supported to the mid-depth of the deck specimen over the simple supports. Figure 4.4 shows

42

these LVDTs being used during an elastic test. Before the test was conducted, all

instrumentation used in the elastic test was calibrated. For simplicity, the data was not

recorded digitally in a file, instead the deflections were read during real-time loading and

each individual deflection was recorded by hand in a table within the project lab book. No

strain instrumentation was used in the elastic test.

CL 9' Width

8

LC

A6A36A24A12

A18

6

7

12"

12"5

LA6, A12,... denotes panel and inches from C

Contact Area of Load

3

12" 12"4

3, 4,... denotes LVDT channel of labview output

(Varies)Concrete Panel Span

Sim

ple

Supp

ort

Sim

ple

Supp

ort

A18 denotes 18" from LC

Figure 4.7 Elastic test pattern for LVDT displacements of a concrete deck panel quadrant

The final testing step for the concrete deck panels was the inelastic loading test. For

instrumentation concrete and FRP strain gauges were applied on top and bottom surfaces

along the centerline of the span to measure strains at various increments from the load patch

43

depicted in Figures 4.8 and 4.9. These strains helped to determine the lateral distribution of

the load as the load was applied. In addition to strain gauges, LVDTs with a ± 1.0” range

were used to measure deflections of the concrete deck panel on the top surface at various

intervals along the centerline. These deflections will provide a centerline deflection profile

as the load was increased to failure. The LVDT locations are depicted in Figure 4.8. The 4”

potentiometer was used to measure the deflection near the load patch to provide a deflection

data for at least one point near the load in case LVDT14 (see Figure 4.8) exceed its deflection

range. The data produced by the instrumentation for the inelastic test was digitally recorded

to a file by the LabView program. It should be noted that for all tests conducted in this

thesis, the calibration factors for all instruments were input into the LabView software before

any data was digitally recorded in a file. Therefore, the calibration factors are reflected in the

recorded data.

44

CL 9' Width

12"

LCControl

LVDT17

LVDT16

LVDT14

LVDT15

12" SG4

0'-2" 12"

12"

SG7

9"

SG7, SG4 denotes strain gauge

LVDT14, LVDT15,... denotes LVDT

4" Potentiometer12"

6"

(11'-6", nts) Panel A Span

Sim

ple

Supp

ort

Sim

ple

Supp

ort

Figure 4.8 Instrument type and location on the top surface of Panel A for the inelastic test

45

C 9' Width

SG was applied to FRP,but not used in test

CLControl

1'-5

"

10"

SG6

SG5

L SG2

SG3

9"

1-1/2"


6"

(11'-6", nts) Panel A Span

Shiplap joint between the FRP deck forms

Figure 4.9 Instrument type and location on the bottom surface of Panel A for the inelastic test

Table 4.2 connects the instruments illustrated above to the channels of the software

program LabView, which was used to digitally record the data. This is the data record table

for the ultimate test of concrete deck Panel A. The data was recorded on the computer as a

text file and this table was essential for converting the text file data to useful information.

Panel B and Panel C data record tables can be found with the illustrations for the instrument

types and locations in Appendix C.

46

Table 4.2 Data record table for Panel A Ultimate Test (2/26/01)

Instrument Type

LabView Channel

Calibration Constant

Comments

Load 1 -40.0 Stroke (±10”) 2 -2.0 4” Potentiometer

3 0.615 12” from load

Not used 4 ±1” LVDT14 5 0.2053 12” from load

±1” LVDT15 6 0.2112 24” from load

±1” LVDT16 7 0.2062 36” from load

±1” LVDT17 8 0.2081 48” from load SG2, FRP 9 0.00285 Under non-epoxy SG3, FRP 10 0.00275 Under epoxy

aggregate SG4, Concrete 11 0.00373 Above SG5 SG5, FRP 12 0.00277 Under epoxy

aggregate SG6, FRP 13 0.00245 Under non-epoxy SG7, Concrete 14 0.00261 Above SG3

Note: The information in the recorded data file reflect the use of calibration constants

4.2.4 Description of Specimen Loading

The following paragraphs will describe how each concrete deck panel was loaded for

each step of the experimental testing. As previously mentioned, each concrete deck panel

was conditioned before any data producing steps were conducted. This was the step that

involved the cyclic loading from 0 to 24 kips. The actuator load was manually controlled

and applied by load control from the MTS 458.20 control unit. After 10 complete cycles

were conducted the conditioning was complete and the concrete deck panel was poised for

the elastic test.

The elastic test was loaded to service load level of an AASHTO HS20-44 wheel load,

16 kip. The actuator load was monotonically and manually applied under load control. Once

47

the instruments were applied and the data acquisition system was recording, the load was

applied from 0 kip to 16 kip in a relatively slow manner. At 16 kips, the load was held

constant, until the service load deflections from the LVDTs were manually read and

recorded. Then the load was removed. Subsequently, the line of LVDTs was repositioned to

the next line of displacement readings. Because of the lack of LVDT instruments, compared

to the number of service load deflection points, this process had to be repeated. Each line of

data points parallel to the span centerline, see Figure 4.8 (centerline, A6, A12…), required a

loading step and data recording cycle. When the last line of displacements, closest to the

support, was read then the load was removed and the elastic test was complete. At this point,

instruments were replaced and added to prepare for the ultimate test step. The LVDT

deflections recorded from each line of data points in the elastic test were normalized for

relative comparison.

The final experimental step, the ultimate test, was manually controlled by both load

and stroke control. The ultimate load test required no prescribed load level. Each concrete

deck panel was loaded until it was determined that it structurally failed. At first, the loading

was controlled by load control. However, once it was judged the specimen had reached it’s

yield or peak load carrying capacity, the actuator control was switched to displacement or

stroke control. Switching to stroke control was necessary to avoid a sudden and catastrophic

failure. In order to record the failure in detail, the failure was slowed and manually

controlled, which can only be carefully done under actuator stroke control. After the ultimate

load capacity was reached and the specimen started to significantly deflect (usually between

a 2”-4” deflection) without maintaining a monotonic load level, it was deemed that the test

was complete.

48

4.3 Positive Moment Beam Test

To characterize and isolate the positive moment behavior of the FRP reinforcement

system in the primary direction, through 3-point bending, a series of three beam tests (Beam

D) were conducted. However, one of three beams was cast with concrete without the top bi-

directional grid. The purpose of this was to evaluate the effect the top grid has upon the

capacity and behavior of the FRP reinforcement system under positive moment forces. Even

though evaluating the impact of the top grid was important, the main focus of the positive

moment beam test was to observe how the FRP deck panel performs as bottom reinforcing in

composite action with the concrete. Each positive moment beam was 10’-10” long, 3’-0”

wide and 8” thick. As seen in Figure 4.10, each beam was subjected to a 3” wide line load

applied at the center of the 8’-8” span.

3'

8'-8"Span

Area of line load,3" wide

LC

10'-10" Figure 4.10 Plan view of a positive moment beam, Beam D

4.3.1 Positive Moment Beam Test Station

The same reaction frame set-up used in the concrete panel test, section 4.2.1, was

utilized for testing the positive moment beam. Supporting the positive moment beam were

much smaller reaction blocks. The dimensions of these blocks were 24” high, 24” deep and

49

39” wide and were formed and poured at SMTL. Again, the reaction blocks were set up on a

layer of plaster to ensure the block evenly distributes the reaction force to the floor. A 5-ton

overhead crane was used to position the positive moment beam on the concrete support

blocks. Just like the concrete deck panel, the positive moment beam was founded on 8” wide

x ¼” to ½” thick plaster bearings. Like the plaster between the support blocks and the floor

these bearings were cast to evenly distribute support for the deck. The bearing also provided

clearance for rotation of the concrete deck panel during testing. As seen in Figure 4.11, the

positive moment beam and supporting assembly was centered in both directions of the testing

frame. This ensured the actuator vectorial load was centered on the 3” square bar used to

transfer the load from the actuator base plate to the top of the specimen. Between the 3”

square bar and the specimen concrete surface was a thin layer of Hydrocal plaster to provide

uniform bearing and to prevent a premature bearing failure by crushing the concrete. The 3”-

square bar line load application assembly is shown in Figure 4.12

50

Figure 4.11 Positive moment beam test set-up

Figure 4.12 Three inch square bar line load assembly

51

A 200 kip capacity MTS closed-loop, servo-hydraulic actuator provided loading to the

positive moment beam. The 200 kip hydraulic actuator can be seen in Figure 4.11, mounted

in the center of the beam of the loading frame. An MTS 458.20 digital servo control station

directly controls the 200 kip hydraulic actuator.


The same data acquisition system and the same type of instrumentation that was used

for the concrete deck panel testing, defined in section 4.2.2, was utilized for the positive

moment beam testing. Hence, the same calibration procedures were applied. Compared to

the concrete deck panel test, however, there was less instrumentation required in a positive

moment beam test.


Unlike the concrete deck panel test, testing of the positive moment beam did not

require a conditioning step. Needing to isolate the positive moment behavior of the FRP

reinforcement system as a whole, testing of the positive moment beam only had two steps:

elastic and inelastic. By conducting the elastic test, it naturally conditioned the beam. These

testing steps are analogous to the elastic and ultimate testing steps of the concrete deck

panels. The paragraphs and illustrations to follow are to document the kind of and location

of instrumentation, as well as, how the data was recorded for each step.

The elastic step was conducted first, here the specimen was loaded manually from 0

kips to 24 kips two times to condition the specimen to a cracked section and data was

recorded. An LVDT, potentiometer and some strain gauges were placed to measure the

deflections and strains on the top concrete surface in addition to strains on the bottom FRP

52

surface. A ±0.1” LVDT was mounted near the centerline of the span on one side of the

actuator to accurately measure the stiffness behavior for the initial deflections. A 4”

potentiometer was symmetrically positioned to the LVDT to monitor the deflections beyond

the range of the ±0.1” LVDT.

Measuring strains near the centerline provided a moment-curvature relationship,

which was a valuable piece of information for an overall evaluation of structural behavior of

the FRP reinforcement system. Therefore, two sets of strain gauges were placed as near to

the centerline of the span as possible. Each set of strain gauges were positioned in the same

position, except one gauge was applied on the top concrete surface, while the other shadowed

the same location on the bottom FRP surface. Both sets were located the same distance from

the centerline of the span, for a bit of redundancy. The same instrument set-up was used for

the elastic and inelastic tests of Beam D1, D2 and D3. However, Beam D1 and D3 had

additional instruments added, including a Soiltest CT-171 Multi-Position Mechanical Strain

Gauge (4” gauge setting) to verify the electronic strain gauge data collected from Strain

Gauge 3. Figure 4.13 and Figure 4.14 illustrate instruments used and location for the top and

bottom surfaces (Beam D1 and D2), respectively. The illustration of Beam D3 instrument

type and location is documented in Appendix C.

53

1'-6

"SG3

7"

SG2

4"

LVDT13

LC Control


4"

8'-8"Span

7" 3"

3'

4" Potentiometer

Figure 4.13 Instrument type and location for top surface of Beam D1 and D2

(Except D1 has a mechanical strain gauge located opposite SG3)

SG3SG34"

SG2

L ControlC

1'-5" 4"

3'

Span8'-8"

FRP deck form

FRP deck form

Figure 4.14 Instrument type and location for bottom surface of Beam D1 and D2

Table 4.3 documents the record of instrument data to the channels of the software

program LabView, which was used to digitally record the data of both the elastic and

inelastic tests. This was the data record table for the ultimate test of both Beam D1 and

Beam D2. The data, which include the calibration effects, was recorded on the computer as a

54

text file and this table was essential for converting the text file data to useful information.

The data record table for Beam D3 can also be found in Appendix C.

Table 4.3 Data record table for Beam D1 (4/10/01) Beam D2 (4/6/01)

Calibration Constant Beam D1


Instrument Type

LabView Channel

Elastic Inelastic Elastic Inelastic

Comments

Load 1 5.0 40.0 5.0 40.0 50k/200k Stroke 2 0.2 2.0 0.2 2.0 2”/10” 4” Potentiometer

3 -0.610 -0.610 -0.610 -0.610

±0.1” LVDT13

4 0.0245 0.0245 0.0200 0.0200

SG2, Concrete 9 3.9636 3.9636 4.240 4.240 Above SG4 SG3, Concrete 10 2.7087 2.7087 2.720 2.720 Above SG5 SG4, FRP 11 8.7846 8.7846 8.790 8.790 Below SG2 SG5, FRP 12 2.8206 2.8206 2.800 2.800 Below SG3

Note: SG calibration constants x1000


The following paragraphs will describe how each positive moment beam was loaded

for each step of the experimental testing. The elastic test involved recording the positive

moment stiffness (via load versus deflection relationship) with more accuracy in the elastic

load range. Deflections were recorded with a sensitive LVDT (±0.1” range). This was the

step that involved the cyclic loading from 0 to 24 kips. The actuator loading cycle was

manually controlled by load control from the MTS 458.20 control unit. After two complete

cycles were conducted and all load was removed, the elastic test was complete and the

positive moment beam was poised for the inelastic test.

The inelastic test was loaded monotonically and started with no load applied to the

beam. Then gradually, under manual load control, the load was increased at an constant rate

until it was judged that failure was imminent. At this point, load control was switch to stroke

55

control. As previously mentioned, the purpose of the control switch to stoke was to manually

control the test through the failure of the beam. This ensured that failure occurs slow enough

to allow the data acquisition to record the failure in detail for latter analysis. The test was

considered complete when the mode of failure was evident or when it was judged unsafe to

continue to induce displacement to the beam.

4.4 Negative Moment Beam Test

To characterize and isolate the negative moment behavior of the FRP reinforcement

system, through 2-span continuous bending, a series of three beam tests (Beam E) were

conducted. Originally, there were four negative moment beams to test. However, one of the

four beams, which was cast with concrete without the top bi-directional grid, fractured into

two pieces during delivery (explanation in Chapter 3). The purpose of the negative moment

beam test was to observe and characterize the FRP bi-directional grid’s performance to

reinforce, in tension, the concrete over a support, such as a girder. Each negative moment

beam was 17’-4” long, 3’-0” wide and 8” thick. In addition, as seen in Figure 4.15, each

beam was subjected to a 3” wide line load applied at the center of each of the 8’-0” spans.

56

8'8'

17'-4"LC

Area of line load, 3" wide

LC

3'

Span Span

Supp

ort

Supp

ort

Supp

ort

Figure 4.15 Plan view of negative moment beam (Beam E)

4.4.1 Negative Moment Beam Test Station

Since an actuator was required for each span, two reaction frames were necessary for

the negative moment beam test set-up. The negative moment beam test set-up is shown in

Figure 4.16. In addition to the 200 kip actuator reaction frame used to test the concrete deck

panels and positive moment beams, a 55 kip actuator reaction frame was implemented. This

55 kip reaction frame was made of steel built up members and the columns were directly

bolted into the structural floor. Controlling the 55 kip actuator was an MTS Test Star control

station. The output stroke and load, from the 55 kip actuator were routed to the data

acquisition system that had been previously used. While conducting the structural test of the

negative moment beam, however, the two actuators with two different control stations

required two people to manually staff the controllers.

The same concrete reaction blocks used for the positive moment beams were used to

support the end of the negative moment beams. A 24” high x 39” wide x 20” thick concrete

block provided the middle support between the two spans. As done previously, each support

block had a layer of plaster between it and the floor to ensure even distribution of loads from

the block to the floor. A 5-ton overhead crane combined with mechanical rollers was used to

position the negative moment beam on the concrete support blocks. During the placement of

57

Beam E5, the lifting procedure produced a full width (partial depth of the section too) crack

in the concrete on the top surface right over the centerline of the middle support and beam.

Therefore, Beam E5 was initially a cracked section over the centerline of the middle support,

the negative moment region, before any test load was ever applied. Beams E1 and E3 both

were installed on the concrete supports without developing this cracked section. Just like the

concrete deck panel, the negative moment beam was placed on 8” wide x ¼” to ½” thick

plaster bearings at each support block. The bearing also provided rotational clearance.

The same 3” square bar assembly used for the positive beam test to transfer the load

from the actuator base plate to the specimen surface, described in Section 4.3.1, was used for

both spans of the continuous beam.

Figure 4.16 Negative moment beam test set-up

58



for the concrete deck panel testing, defined in section 4.2.2, was utilized for the negative

moment beam testing. Hence, the same calibration procedures were applied. Again,

compared to the concrete deck panel test there was less instrumentation required in a

negative moment beam test.


The goal with the negative moment beam test was to isolate the negative moment

behavior of the FRP reinforcement system as a whole, particularly measuring strains of the

section over the middle support. As previously mentioned, over a support (such as a girder)

was where the top bi-directional grid reinforced the concrete in tension. The testing of the

negative moment beam only had one step, inelastic. The inelastic testing step conducted on

the negative moment beam was similar to the inelastic testing step of the concrete deck panel

and the positive moment beam. However, here two spans were loaded simultaneously. The

paragraphs and illustrations to follow document the kind of and location of instrumentation,

as well as how the data was recorded for the inelastic testing step.

In addition to the strains measurements of the section over the middle support, the

deflection in each span was recorded. To measure the center deflection in each span, a ±1.0

inch LVDT was mounted near the line loads. Figure 4.17 depicts the location of these

LVDTs in detail.

59

8'8'

LC LC

3'

200kip Actuator Span 55 kip Actuator Span

3'-8"7" LVDTA 7"LVDTB

3'-8"

1-1/4"SG1 or SG2

A

A Ctl

Figure 4.17 Instrument type and location of the top surface of negative moment beam

Measuring strains of the section over the support provided a moment-curvature

relationship, which was a valuable piece of information for an overall evaluation of structural

behavior of the FRP reinforcement system. Figure 4.18 illustrates that over the middle

support the FRP deck form was not continuous. In fact, there was an 8” wide concrete

surface between the bottom FRP deck forms. The center of this 8” width was the centerline

of the beam, bearing, and support. The cut section of Figure 4.19 exhibits the location of

three gauges positioned and projected at the same spot along the depth of the section: one on

the top surface, another on the center in depth (on the web) of the 2” FRP “T” bar, and a

strain gauge sandwiched between the bottom concrete surface of the beam and the top of the

plaster bearing. The profile of these gauges provided three points of strain reference through

the depth of the section. One of the three negative moment beams, Beam E5, was cast

without the strain gauge on the web of the FRP “T” bar and, therefore, only had two strain

gauges through this section.

Cracks that developed over the middle support during the test of Beams E1 and E3

were measured with a magnified ruler and manually recorded in the lab book. The accuracy

60

of the magnified ruler used to monitor the cracks during the test was about a hundredth of an

inch. This rough estimate was necessary to conduct because the concrete strain gauges

applied to the top concrete surface were very close to the cracks that developed over the

middle support early in the test. Once these cracks developed, the close proximity of

concrete strain gauge seriously impaired ability to measure strains with any accuracy.

CL CCtlL

1-1/4"

SG3

3'

FRP Deck Form

8"Concrete Section

Figure 4.18 Instrument type and location of the bottom surface of negative moment beam

8"

1'-6"

CL

5-1/

2"

SG2, Beam E5SG1, Beam E1 & E3

SG2, Beam E1 & E3

SG3, All E Beams Figure 4.19 Cross-section A-A of Figure 4.17, showing strain gauge layout

Table 4.4 documents the record of instrument data to the channels of the software

program LabView, which was used to digitally record the data (with the calibrations effects

61

included) of the inelastic test. This is the data record table for the ultimate test of three

negative moment beams, Beam E1, Beam E3 and Beam E5.

Table 4.4 Data record table for Beam E1 (5/9/01), Beam E3 (5/16/01), Beam E5 (5/2/01)

Instrument Type

LabView Channel


Beam E1 E3 E5 E1 E3 E5

Comments

200k Load 1 1 1 -20 -20 -20 200 kip span 200k Stroke (±10”) 2 2 2 -1.0 -1.0 -1.0 200 kip span ±1” LVDT A 3 3 3 -0.2068 -0.2068 -0.2053 200 kip span ±1” LVDT B 4 4 4 0.2049 0.2049 0.2049 55 kip span 55k Stroke (±3.1”) 5 5 5 -1.0 -1.0 -1.0 55 kip span 55k Load 6 6 6 -10.0 -10.0 -10.0 55 kip span SG1, top of Concrete

9 9 - 3.956 4.207 - No SG1 in Beam E5

SG2, location varies 10 10 9 2.726 2.728 4.05 SG3, bottom of Concrete

11 11 10 8.703 8.722 2.71

Note: The strain gauge calibration constants are x1000


During the inelastic test the negative moment beams were manually loaded

monotonically in each span. Each span of the continuous beam required a separate actuator

and controller; therefore two manual operators were required to apply the loads during the

test. Even though the load and stroke data was recorded for both actuators, the goal was still

to apply the same loading to each span during the test. This was accomplished by applying

the load at a slow rate while one of the actuator controllers, the “master” controller, was

verbally calling out the load being applied. At the start of loading, both actuator controllers

were under load control. The other operator would apply the same load being verbally

communicated by the “master” controller. However, since one of the actuator’s capacity was

55 kips, it reached its limit during the test. At this point, the 55 kip actuator would hold its

62

limit load, 50 kips, while the 200 kip actuator would continue to apply a load to the other

span. The loading continued under load control until failure was imminent. From

observation, when failure was near, control of the 200 kip actuator was switched to stroke

control. As mentioned in previous beam tests, this allowed a slow control of the actuator

through the ultimate region. The test was complete during post-ultimate deflections, of the

200 kip actuator span, when it was judge enough data was recorded for a thorough

interpretation of beam performance or the test was stopped for sake of safety.

4.5 Accelerated Fatigue Beam Test

Part of serviceability requirements for concrete bridge decks deals with fatigue issues

of the materials. Therefore, an accelerated fatigue beam test, Beam E2, was performed to

evaluate the fatigue resistance of the FRP materials of the bi-directional grid and Stay-in-

place deck form. In addition to the fatigue resistance of the material, this FRP reinforcement

system relied upon the composite action between the surface bond of the FRP deck form

epoxied with aggregate and the concrete. Since the structural integrity of the FRP system

relies upon this bond, it too was subjected to fatigue testing. All categories and all materials

for AASHTO design specification fall under a fatigue resistance of 2 million cycles or less,

therefore testing this FRP reinforcement system to the maximum requirement of 2 million

cycles was sufficient.

The geometry of the accelerated beam, Beam E2, was identical to the two-span

continuous negative moment beam, 17’-4” long, 3’-0” wide and 8” thick. Consequently,

Figure 4.15 is applicable for the fatigue beam as well.

63

4.5.1 Accelerated Fatigue Beam Test Station

The test station used for the negative moment beam testing, described in section

4.4.1, was exactly what was used for the accelerated fatigue beam test. There was one

difference between the two test set-ups; there was no requirement for two manual operators

to conduct the test. The test was conducted on “autopilot”, each controller (MTS 458.20 for

the 200 kip actuator and the Test Star for the 55 kip) was programmed to apply set loading

conditions. These loading prescriptions will be explained in detail in section 4.5.4.



for the negative moment beam test, defined in section 4.4.2, were utilized for the accelerated

beam test. Other than load and stroke functions of the two controllers, no other

instrumentation was used.


As just mentioned, there was no LVDT, potentiometer or strain gauges used to

measure any data during this test. The purpose of this test was to measure the relative

stiffness loss over 2 million cycles and this was achieved by periodically measuring the

stroke output of the actuator controllers.

The stiffness data recorded was read from the stroke displacement output of each

controller and manually recorded in lab book tables. The Test Star controller also digitally

recorded a three-minute record of stroke displacements at every 200,000 cycles.

64


The 200 kip actuator control was set to hold a static load of 16 kips in one span. In

the other span, the 55 kip actuator controller was programmed to apply a cyclic load, at two

hertz, with an amplitude of 16 kips ranging from 4.8 kips to 20.8 kips. Both controllers were

set for load control and since the majority of the testing was unmanned, limit switches for

excessive deflections were set to prevent catastrophic failure of the specimen or any

unforeseen accidents. The locations of the reaction block and the specimen were monitored

for any translations of the set up during testing. Any concrete cracking was of the beam was

monitored during the test at 10 kip intervals.

At every 200,000 cycles stroke displacements and load were recorded for each span.

Each series of readings involved setting the load control of one actuator to a static load of 16

kip. The other span was then unloaded and then reloaded at 2 kip increments up to 16 kips.

At each 2 kip increment, the stroke displacement was read. When the last stroke recording

was read, the load was kept at 16 kip while the process was repeated for the other span.

After both spans were tested, the cyclic loading was restarted. The goal was to monitor the

relative stiffness loss over 2 million cycles.

After the 2 million-cycle test was complete, the fatigue beam was loaded to ultimate

in the same manner as the negative moment inelastic test and with similar instrumentation,

see Section 4.4.4. The data record table is shown in Table 4.5 and the inelastic data was

intended for a load versus deflection comparison to the negative moment beams not exposed

to an initial 2 million cycles of service load.

65

Table 4.5 Data record table for the inelastic test of Beam E2 (fatigue beam)

Instrument Type

LabView Channel


55k Load 1 -10 55k Stroke (±3.1”) 2 -1.0 ±1” LVDT A 3 -0.207 ±1” LVDT B 4 0.209 200k Stroke (±10”) 5 -1.0 200k Load 6 -20.0

Note: The information in the recorded data file reflect the use of calibration constants

4.6 An Additional Concrete Panel Test

As the course of the testing was conducted, it became apparent from test results that

the FRP panel reinforcement system was only achieving partial composite action and

significant longitudinal concrete cracking occurred at the location over the shiplap joints.

Therefore, a decision was made to construct an additional concrete deck panel, Panel C2,

with modifications to the FRP deck panel and the inclusion of reinforcement above the

shiplap joints. To measure the increase the strength performance through greater composite

action, additional coverage of epoxyied aggregate was provided on all horizontal surfaces of

the FRP deck form. To see if some reinforcement in the concrete above the shiplap joints

could increase performance, Fibergrate Molded Grating was added. See Figure 3.9 for a

cross sectional view of Panel C2. The additional concrete deck panel had the same geometric

dimensions as Panel C, 8’-0” span x 9’-0” wide x 8” thick.

66

4.6.1 Concrete Deck Panel Test Station

The same test station assembly used for the concrete deck panels in Section 4.2.1 was

applied to this additional concrete deck panel, Panel C2. The plan view of Figure 4.1 is also

valid for Panel C2.


Section 4.2.2 explains in verbatim the data acquisition system and the type of

instrumentation used for Panel C2.


The exact same battery of testing steps (conditioning, elastic and inelastic) explained

in Section 4.2.3 are applicable to testing of Panel C2. The conditioning and the elastic test

(see Figure 4.8) of Panel C2 used the same procedures and instruments as the other concrete

panels. There was a difference in the location of the instruments used in the inelastic test for

Panel C2. Figures 4.20 and 4.21 exhibit the type and location of the data recording

instruments applied on the top and bottom surface, respectively. Also, Panel C2 was

implemented with more strain gauges within the section, applied to the surfaces of the FRP

bi-directional grid and FRP deck form. Figure 4.22 depicts the interior strain gauge

locations.

67

C 9' Width

12"

ControlCL

LVDT17

LVDT16

A

12"

LVDT15

A

24" 14-1

/4"

SG2

L



4" Potentiometer 12"SG17

9"

4"

(8'-0", nts) Panel C2 Span

Sim

ple

Supp

ort

Sim

ple

Supp

ort

Figure 4.20 Instrument type and location for top surface of concrete deck Panel C2

(inelastic test)

68

C 9' Width

CLControl

L

L

14-1

/4"

SG6

C

LVDT13, monitor horiz. joint opening 9"

SG7


4"

(8'-0", nts) Panel C2 Span


Figure 4.21 Instrument type and location for bottom surface of concrete deck Panel C2

SG2

SG6SG5

SG3

SG5

1-1/2"

3-1/

2"

2-1/2"

0'-8

"

Figure 4.22 Profile strain gauge location of section A-A of Figure 4.20

(Fibergrate Molded Grating reinforcement above shiplap joint not shown)

Table 4.6 connects the instruments illustrated to the channels of the software program

LabView, which was used to digitally record the data (data reflects calibration effects). This

69

is the data record table used to link the information recorded in the data file to the

instruments used for the ultimate test of concrete deck Panel C2.

Table 4.6 Data record table for Panel C2 Ultimate Test (6/27/01)

Instrument Type

LabView Channel


Comments

Load 1 -40.0 200 kip Stroke (±10”) 2 -2.0 4” Potentiometer 3 6.20 Cal. Const x10 ±0.1” LVDT13 4 -1.994 Cal. Const x100 ±1” LVDT15 5 2.07 Cal. Const x10

±1” LVDT16 6 2.12 Cal. Const x10

±1” LVDT17 7 -2.10 Cal. Const x10

SG17, Concrete 8 -3.91 Top of SG7 SG2, Concrete 9 -3.455 Top of 5 SG3, FRP 10 -4.04 Top flange of I bar SG4, FRP 11 -3.97 Bottom flange of I SG5, FRP 12 -3.92 Top of corrugation SG6, FRP 13 -4.02 Bottom of section SG7, FRP 14 -3.97 Bottom of SG17

Note: all SG calibration constants are x1000


The loading steps and the loading sequence described in Section 4.2.4 were used for

the modified deck panel test. The conditioning step and the elastic step went according to the

plan spelled out for the other concrete deck panels. The ultimate capacity test, however, was

cut short of achieving an ultimate failure. The reaction frame of the 200 kip actuator started

to lift off the structural floor when the load approached an ultimate value of around 120 kips.

As a result, the ultimate or inelastic test was cut short of the goal of monitoring a post-

ultimate behavior of the modified concrete deck panel.

70

5 TEST RESULTS

5.1 Introduction

The description that follows was organized by tests that were conducted for a similar

purpose. In part, the main topics for each section will describe the following: how the test

was conducted, visual information obtained during the test, data measured during the test,

when the test was halted and judgment of failure. The broad aspects of how each test was

conducted will be recalled, but the details of this testing can be found in Chapter 4. This

chapter will focus on the inelastic test results for each specimen, except for the fatigue test.

Not all data the from all the tests conducted will be presented, however, when test data from

either the basic conditioning or elastic steps can help address the objectives, then this

information was included. Finally, it should be noted that all test data does not include

selfweight effects.

5.2 Concrete Deck Panel Test

In general, the concrete deck panel test was designed to answer three key unknowns

of the proposed FRP reinforcement system: how does the lack of bottom distribution steel

and presence of shiplap joints between the FRP deck forms affect the distribution of the load,

the mode of failure and the ultimate capacity. Each concrete deck panel was subjected to

sequential testing steps of conditioning, elastic and inelastic testing. The methodologies of

conditioning, elastic and inelastic testing steps were described in Section 4.2. For the

concrete deck panel, the results obtained from the elastic tests were recorded, however, at

this point this information was not directly essential to address the research objectives. In

71

situations where these results can help meet the objectives, they will be incorporated into the

test results. On the other hand, inelastic testing, which includes testing through the elastic

region, naturally produced informative data requisite for developing load distribution

characteristics, determining failure mechanism and establishing strength.

Inelastic testing involved loading each specimen over a centered surface area

approximately equivalent to the contact area of a double tire wheel, see Figure 4.1. The 16th

edition of AASHTO LFD design specification defines the contact area as a function of the

wheel load. Since the wheel loading in the inelastic testing phase was not static and the

contact area remained constant, the area shown in Figure 4.1 was judged sufficient. The

monotonic load was applied under manual load control at a relatively slow rate. While

loading, deflections and material strains were recorded. In addition to the digital data, visual

and audio observations were also recorded. For example, all preloading concrete cracks were

traced in marker and noted. Near the ultimate capacity point, the actuator was switch from

load to stroke control and loading, indirectly through displacement control, continued. Each

concrete deck panel was loaded past ultimate capacity, except for Panel C2, until a

satisfactory level of displacement was reached. The test was terminated when a significant

deflection beyond the ultimate capacity was safely achieved.

5.2.1 Panel A, 11’-6” Span

Being the first test conducted of the entire project, audio and visual details were

somewhat neglected. As can be seen in Figure 5.1, all concrete surface cracks that existed

before loads were applied were traced and noted with a pen marker. There were a handful of

these preload size surface concrete cracks for Panel A that had little consequence on, or

72

contribution to, the structural performance. As the test progressed, all new concrete cracks

and continued growth of existing concrete cracks were traced and the load at a specific crack

length was noted. In the end, the marked concrete cracks helped to illustrate the values of

tension forces and how they progressed as load was applied.

Figure 5.1 Preload concrete surface cracks at one supported end of Panel A

LabView software was used to record the data from all of the instruments used. This

software allowed a live, real-time, view of a load versus mid-span deflection relationship, via

a graph, while the load was applied to the specimen. The deflection was monitored at a point

12” offset from the center of the load patch, along the span centerline of the concrete deck

panel (4” Potentiometer). Hard data recorded during the inelastic test pertained to vertical

deflections and material strains at locations, on Panel A, depicted in Chapter 4.

Not surprisingly, the initial run experienced a few errors. The data acquisition system

experienced an operator mistake and data was not recorded for any instrument input data for

the first 36 kips of load. The loss of this instrument data inhibited any data analysis within

the initial elastic load versus deflection region. Although small, this undoubtedly introduced

an unknown error into the data, because the initial offsets at no load are not known and the

73

data cannot be properly zeroed. Another error was neglecting to document the audio

information, such as peculiar cracking noises, as the test was conducted. Although it can be

easily overlooked, some audio information can provide explanations to structural behavior.

Adjustments were made and both problems were resolved before the next concrete deck

panel test was performed.

0

10

20

30

40

50

60

70

80

90

100

0.0 0.5 1.0 1.5 2.0 2.5 3.0 3.5 4.0 4.5

Stroke, inch

Lo

ad, k

ips

P yield = 80k

K effective= 88.1 k/inch

Figure 5.2 Load versus actuator stroke for the inelastic test of Panel A

The profile of the load versus actuator stroke response of the inelastic test is

illustrated in Figure 5.2. As will be discussed, since the potentiometer original offsets were

unknown and the uplift of the reference frame near the ultimate load, the stroke data was

used rather than the potentiometer deflection data. There seemed to be a region of softening

just before failure that maintained a high load resistance through increasing deflection.

Although service deflections are not apart of the scope of this thesis, it should be noted that

the interpolated stroke deflection at 20 kip (approximately HS20-44 with impact) was 0.198”,

which equals L/700. This stroke data includes some small deflection contributed by

74

compression of the elastomeric pad under the actuator load. The actual deck deflections,

which did not include these extraneous deformations, would show stiffer behavior. The

stroke versus deflection relationship can be defined as follows: an initial elastic region of

high stiffness, followed by a slight decrease in stiffness, then ultimate capacity, and finally

followed up by a post-ultimate softening/degrading region.

At a stroke defection of 1.70”, the ultimate capacity of 96 kips was achieved. Soon

after reaching the ultimate capacity, each of the four corners of the deck significantly lifted

(easily observable) off their respective support blocks, which is classical behavior for a

simply supported slab. The lifting of the corners can introduce a disagreement between

stroke data and relative displacement from instruments mounted on the reference frame,

because the reference frame was directly attached to each corner. So, the difference between

the stroke (at the center) and the 4” potentiometer (12” offset from center) readings was

graphed in Figure 5.3 to determine if the lifting of the four corners introduced error into the

potentiometer and the LVDTs. There was a constant difference, relatively, until 1.61” or just

before ultimate. This constant difference indicates that lifting of the corners did not adversely

affect the displacement measurements until just before ultimate capacity was reached. After

1.61” the difference between the stroke and the potentiometer sharply increases at a constant

rate because of the onset of punching shear. Therefore, only the stroke values were valid.

The post-ultimate test was terminated at a stroke deflection of 4.10” and still carrying 63

kips. Although difficult to visually define, a yield point of 80 kip and an initial stiffness of

88 kips per inch were characterized by a standard technique proposed by Preistley (1992),

which will be used for all specimens to follow for relative comparisons. The Preistley

method is explained in detail in Appendix D.

75

0

20

40

60

80

100

120

0.0 0.5 1.0 1.5 2.0 2.5 3.0 3.5

Difference between stroke and potentiometer, in.

Load

, kip

Figure 5.3 Difference between the stroke and potentiometer readings, Panel A

There was no concrete cracking along the unsupported sides of the panel, leading to a

conclusion that there was little load distributed all the way out to the unsupported edges. The

ends of the panel that rested on the reaction blocks did see some vertical concrete cracking,

especially in the middle region of each end. Some of these vertical cracks along the

supported ends were connected to the oval surface cracks seen in Figure 5.4.

76

SUPPORT

SUPPORT

Figure 5.4 Oval concrete cracking pattern, post-inelastic test of Panel A

Figure 5.5, shows the top surface deflections across the centerline section at various

load levels. The locations of the potentiometer and the LVDTs used to measured these

deflections are illustrated in Figure 4.8

77

-1.8-1.6-1.4-1.2-1.0-0.8-0.6-0.4-0.20.0

-54 -36 -18 0 18 36 54

Distance from actuator, in

Def

lect

ion

, in

36.4 kips 40.5 kips 50.2 kips 61.2 kips70 kips 80 kips 90 kips 96.5 kips (Ult.)

Edge EdgeCenterline

Figure 5.5 Deflection profile of Panel A at the center span

In addition to measuring deflections, material strains were monitored and recorded.

Figure 5.6 shows the strains measured on the top and bottom surfaces of the concrete and

FRP, respectively, for various load increments. Illustration of the strain gauge locations on

Panel A are documented in Figures 4.8 and 4.9. It was clear that the FRP strain gauge closest

to the actuator, SG2, did not read any strains and apparently was not functional. Other than

this errant data, the strain readings as a whole are reasonable and follow the deflection trend

in Figure 5.5. The trends of the strains are logically increasing as the load increases and

decrease as the distance from the load increases. It must be kept in mind, however, that

because the instrumental data recording did not start until 36 kips into the test, the zero

offsets for all the instruments, except load, are unknown. This invariably introduces some

small, and probably insignificant, unknown error.

Other point of interest was the magnitude of the strains seen in the material. Near the

ultimate load the concrete and the FRP strains are greater than 0.5% and 1%, respectively.

78

These strains seem a little higher than what would be anticipated, especially the concrete

strains (>0.003). Concrete is not a linear-elastic material and the concrete modulus of

elasticity decreases dramatically near ultimate and therefore the strains increase at a non-

linear rate. In addition, the higher concrete strains can be attributed to a state of stress where

the perpendicular confining effect allows higher strains to be reached before crushing.

-11

-9

-7

-5

-3

-1

1

3

5

-54 -36 -18 0 18 36 54

Distance from the actuator along the center of span, inches

Stra

in, m

icro

stra

in

36.4 kips 59.1 kips 70 kips 80 kips 96.5 (Ult.)

PanelEdge

PanelEdge

Compression

Tension

Concrete

FRP

Figure 5.6 Longitudinal strains recorded along the centerline of span, Panel A

From the test results for Panel A, three main observations or reasons lead to the

conclusion that the mode of failure for Panel A was punching shear. First, the top surface

oval cracking pattern seen in Figure 5.4 is consistent with punching failure and inconsistent

with flexural failure where the top would be expected to be in compression. Second, the

concrete crushing occurs only beneath the loading patch. A third indicator was the

differential deflection of the two middle FRP deck forms (located directly under the load

patch) from adjacent deck forms. As the actuator was driven down through the panel after

the ultimate load was reached, the middle two FRP deck forms vertically deflected

79

downward from the other FRP deck forms a significant amount, as seen in Figure 5.7. This

differential deflection indicates that there must have been vertical shear cracking at the

shiplap joints.

Panel Separation

Edge of Deck Form Under Load

Edge of Outside Deck Form

Figure 5.7 Differential deflection of the middle two FRP deck forms

5.2.2 Panel B, 9’-10” Span

The test procedure for inelastic testing was almost the same for all of the concrete

deck panels. This time, testing of Panel B, there was visual and audio information noted as

the test progressed. Learning from Panel A, audio information was a necessary additional

source of information that was noted and recorded in the lab book. Also learned from Panel

A was the need to monitor the horizontal opening of the center shiplap joint directly below

the load with a ±0.1” LVDT (see Appendix C for exact location).

Any surface concrete cracks were once more noted before the inelastic test was set in

motion. Again, the surface concrete cracking pattern was marked as the cracks grew with

their respective loads values. LabView software displayed, in graph form, the real-time load

versus mid-span deflection response for the same location as was monitored for Panel A.

Hard data recorded during the inelastic test pertained to horizontal and vertical deflections,

plus the material strains at locations on Panel B depicted in Appendix C.

80

010

20

30

4050

60

70

8090

100

0 0.5 1 1.5 2 2.5 3

Adjusted Stroke, in.

Lo

ad, k

ips

P yield = 77.4 kip

K effective = 152 k/in

Figure 5.8 Load versus adjusted actuator stroke for the inelastic test of Panel B

The load versus deflection profile seen in Figure 5.8 has a similar pre-ultimate region

behavior as seen in Panel A. Again, the load versus deflection region just prior to the

ultimate showed softening, carrying a large percentage of the ultimate value through large

deflections. Usually at every 10 kips the test was briefly paused to mark concrete cracks.

Since this part of the test (elastic region) was under load control, at these times of pause,

deflections would continue to grow under constant load. Therefore, during the data

processing, the stroke data was adjusted to eliminate the static load crack growths in order to

present a corrected load versus deflection profile seen in Figure 5.8. The ultimate load was

87 kips, which occurred at a stroke of 0.88”. Just before the ultimate load was reached, the

load region between 60 and 80 kips produced a crescendo of cracking of the concrete on the

bottom side of the panel (at about 70 kips) followed by sharp popping noises (that were

assumed to come from the FRP deck forms and may have been caused by some de-bonding).

At the same time, many small pieces of concrete started to fall from the shiplap joints to the

81

floor below. The softening noted in the plot was apparently the result of considerable

concrete cracking and possibly some de-bonding of the FRP deck forms. Like Panel A, soon

after reaching the ultimate capacity each of the four corners significantly lifted, as seen in

Figure 5.9, off their respective support blocks.

Vertical Crack

Figure 5.9 View across one support showing uplift of one of the four corners and a large

vertical crack during Panel B test

Again, the difference between the stroke and potentiometer was graphed to examine

if the panel corners lifting off the support blocks caused any difference in the displacement

measurements. Figure 5.10 reveals evidence that the negative difference between the stroke

and the potentiometer near ultimate load was due to the corners lifting the reference frame.

The large vertical crack in the left picture was apparently from punching shear that

prorogated to the supported edge (not seen in Panel A). The stroke deflection at 20 kips was

0.103”, which equals L/1262. Finally, the initial stiffness based on stroke was calculated at

152 kips per inch and the yield was estimated to occur at 77 kips.

82

0102030405060708090

100

-0.5 0 0.5 1 1.5 2

Difference between stroke - potentiometer data, in

Lo

ad, k

ips

Figure 5.10 Difference between the stroke and potentiometer readings for Panel B

The post-ultimate test was carried on through an adjusted stroke deflection of 2.74”

(adjusted to remove static load deflections introduced during pauses of test) with resistance

staying above 70 kips. Since, termination of this test was somewhat based on safety, the

deflection at the termination of the test was smaller than that of Panel A. Again, there was no

concrete cracking along the unsupported sides of the panel, leading to the conclusion that

there was little load distributed to the unsupported edges. As mentioned, the supported ends

of the panel did see some large vertical shear concrete cracks, see Figure 5.9, especially in

the middle regions. Some of the smaller vertical cracks along the supported sides were

connected to the oval top surface cracks seen in Figure 5.11.

83

Figure 5.11 Plan view of the oval concrete crack pattern on the top surface of Panel B

-1

-0.9

-0.8

-0.7

-0.6

-0.5

-0.4

-0.3

-0.2

-0.1

0-54 -36 -18 0 18 36 54


Ad

just

ed d

efle

ctio

n,

in

20 kip 40 kip 60 kip 70 kip 80 kip 87.4 kip (Ult.)

Edge EdgeCenterline

Figure 5.12 Deflection profile of Panel B along the span centerline

Another plot of the deflections, Figure 5.12, shows the deflections across the center

span section at various load levels. The locations of the potentiometer and the LVDTs used

to measure the deflections are illustrated in the Appendix C. The displacement values plotted

84

at the centerline were from the actuator stroke while the other values were from the LVDTs

and the Potentiometer. The difference between the stroke at center and the adjacent

potentiometer reading 12” away at ultimate were due to the different reference frames. The

stroke was relative to the actuator support frame, the LVDTs and Potentiometer were relative

to the deck corners. Since the corners uplifted, the LVDTs and Potentiometer values were

larger than the stoke value as the ultimate load was approached (see Figure 5.10).

The material strains, for set load intervals, along the center of the span are seen in

Figure 5.13. The locations of the strain gauges are illustrated in Appendix C. There were

problems with the number of significant digits recorded for each of the strain gauge channels.

There were only three significant digits recorded, therefore, the data was a relatively rough

estimate, since the recorded accuracy of the strains was ±0.001. This problem was mitigated

for the test specimens to follow by multiplying the calibration constants in the acquisition

software by a known magnitude (multiply by 100 or 1000).

The strains values of the centerline strain gauge of the FRP deck form, strain gauge 2,

show a drop from the adjacent strain gauge. This result seems counter intuitive and this

discrepancy may be due to the inaccuracy of the data previously mentioned or it may be due

to the fact the strain gauge was located below an area without the epoxy coated aggregate,

thus measuring the strains of a shear lag.

85

-10

-8

-6

-4

-2

0

2

4

6

-54 -36 -18 0 18 36 54

Distance from actuator across the center of span, inches

Str

ain

, mic

rost

rain

30 kip 50 kip 60 kip 70 kip 87.4 kip (Ult.)

PanelEdge

PanelEdge

Tension

CompressionConcrete

FRP

Figure 5.13 Material strains recorded along the centerline of span, Panel B

As previously stated in this section, the differential horizontal deformation across the

center shiplap joint was also recorded. This joint opening represents the horizontal

separation of the two middle FRP deck forms as the load was applied. Basically, it can

indicate if there was longitudinal concrete cracking along the shiplap joint. Without

distribution reinforcement in this FRP reinforcement system, concrete cracking along the

shiplap joints may become a serviceability problem. Figure 5.14 shows the load versus

deformation curve of the LVDT used to measure this lateral deflection. The LVDT range

was exceeded just before 70 kip, so any data after this load was not applicable. It can be

noted that a significant change in stiffness occurred at about 28 kips and the rate of joint

opening increased. This must indicate the start longitudinal concrete cracking along the

shiplap joint. The initial longitudinal concrete cracking at the shiplap joint may have been

initiated during the conditioning step, however, where the panel was subjected to a load of 24

kips. If longitudinal cracking started in the conditioning step, then the longitudinal cracking

was initiated below a load of 28 kips.

86

0

10

20

30

40

50

60

70

80

0 0.02 0.04 0.06 0.08 0.1 0.12 0.14

Lateral opening of shiplap joint, inch

Lo

ad, k

ips

K = 2021k/inch

stiffness change at 28k

Figure 5.14 Load versus lateral displacement opening of the center shiplap joint of Panel B

Figure 5.15 gives a visual picture depicting the post-testing lateral shiplap joint

opening between the two middle FRP deck forms at the end of the test.

Figure 5.15 Lateral opening of the center shiplap joint for Panel B

As mentioned earlier, an oval concrete cracking pattern developed on the top surface

of Panel B. The middle two FRP deck forms deflected in unison and separated themselves

from the exterior two FRP deck forms on each side as exhibited by the cracking of Figures

5.7 and 5.15. This evidence combined with the fact that the ultimate capacity of Panel B,

87

with a 9’-10” span, was relatively equal to (actually 10% less than) the ultimate capacity of

Panel A, with an 11’-6” span, leads to a conclusion of a punching shear failure in both tests

since the flexural moment for a given load would have been higher in Panel A (longer span)

than Panel B.

5.2.3 Panel C, 8’-0” Span

For Panel C, there were a few noteworthy issues pertaining to the conditioning step.

During the conditioning step (ten-cycles from 0 to 24 kip), the first cycle was accidentally

taken above the prescribed 24 kip limit to a 30 kip load. The first cracking was heard at a

19.5 kip load of the first cycle of the conditioning step and cracking sounds continued

periodically up to an errant 30 kip loading, 6 kips above the target. The source of this

cracking was unknown. No cracking was detected for the remaining nine cycles to follow.

Again, LabView provided a load versus mid-span stroke deflection real-time graph

for the controller to monitor during the inelastic test. Hard data recorded during the inelastic

test pertained to vertical and horizontal (to monitor shiplap joint) deflections and material

strains at locations on Panel C depicted in Appendix C.

88

01020

3040

5060

708090

100

0 0.5 1 1.5 2 2.5 3 3.5

Stroke deflection, in

Lo

ad, k

ips

P yield = 80 kip

K effective = 153 kip/in

Figure 5.16 Load versus actuator stroke for the inelastic test of Panel C

Figure 5.16 shows the stroke versus the deflection data. The stroke deflection at 20

kips was 0.083”, which equals L/1157. In addition to noting the deflection measurements,

audio information recorded during the inelastic test indicated that the first large crack

probably occurred at a load of 37 kips. The difference between the stroke and the

potentiometer deflection measurements are plotted in Figure 5.17. This graph again provides

substantiation of the visually noted uplift of the four corners of the panel when the load

approached the ultimate capacity. A substantial, and continuous, cracking and popping noise

started at 80 kips, which happened to be the calculated yield (standard technique of Preistley,

1992), and continued up through the ultimate load. The deck also exhibited softening, Figure

5.16, while the noises were noted. The ultimate load occurred at 92 kips with a stroke

deflection of 1.06”. The post-ultimate testing continued on to a deflection of 3” when it was

terminated for safety concerns, but still supporting 62 kip. Finally, initial stroke stiffness was

calculated as 153 kips per inch.

89

0102030405060708090

100

-0.4 -0.2 0 0.2 0.4 0.6 0.8 1 1.2 1.4Difference between stroke and potentiometer, in

Lo

ad, k

ips

Figure 5.17 Difference between the stroke and potentiometer readings for Panel C

Figure 5.18 View of the oval concrete pattern on the top surface of Panel C

(supports are top and bottom of photo)

As seen in Figure 5.18 the top surface concrete cracking pattern was again oval,

somewhat symmetrical in nature, and closely reflects the patterns observed in Panels A and

B. Once again, there was no concrete cracking along the unsupported sides of the panel,

90

leading to the conclusion that there was little load distributed to these edges. The supported

ends of the panel did see some vertical concrete cracking. There were some very large

vertical shear cracks on the ends of the panel, like the ones seen in Panel B (Figure 5.9).

Some of these vertical shear concrete cracks along the supported ends were due to

longitudinal cracking along the shiplap joint and were connected to the oval surface cracks.

As in the previous concrete panel tests, the two middle FRP deck forms, on the

bottom surface of the deck panel, differentially deflected below the adjacent set of two panels

on each side. As seen in Figure 5.7, Figure 5.19 exposes the degree of differential deflection

between these panels.

Differential Form Deflection

Figure 5.19 Elevation view of the bottom surface of FRP deck forms deflecting apart

91

-1.4

-1.2

-1

-0.8

-0.6

-0.4

-0.2

0

-54 -36 -18 0 18 36 54


Def

lect

ion

, in

20 kip 40 kip 60 kip 80 kip 92 kip (Ult.)

Edge EdgeCenterline

Figure 5.20 Deflection profile of Panel C across the center of span

Figure 5.20 depicts the deflection of the center of the span at various load increments.

The corners lifting off the reaction blocks (in the process lifting the reference frame with it)

affected the deflection profile near ultimate relative to the stroke value used to plot the

centerline deflection. The movement of the reference frame was proved by the stoke

deflection being less than the adjacent potentiometer reading (attached to the reference

frame) near the ultimate load.

As done in the previous concrete panel tests, material strains were measured and

recorded along a section at center span. Figure 5.21 illustrates the strain profile along that

section for both the top concrete surface and bottom FRP surface. Counter intuitively the

strain gauge centered below the actuator (SG2) read strain values below a strain gauge

located 9” away (SG3). This defies the conventional thought that the strains would reduce as

distance from the load increases. This discrepancy could be attributed to three possibilities.

First, the strain gauge was not functioning properly through ineffective assembly or

inaccurate calibration. Or this discrepancy may be due to the fact the center strain gauge was

92

below a region of the FRP deck form without epoxied aggregate. This put the gauge in an

area of strain that lags behind areas that have epoxy aggregate to transfer horizontal shear

forces between the concrete and the FRP deck form. Finally, this lag in strain may be due to

the shiplap joint, centered under the load, deflecting enough to detach the surface between

the concrete and the deck form under the width of the steel plate providing the load patch.

-6

-4

-2

0

2

4

6

-54 -36 -18 0 18 36 54

Distance from actuator along centerline of span, in

Str

ain

, mic

rost

rain


Edge Edge

Compression

TensionFRP

Concrete

Figure 5.21 Material strains recorded along the center of span, Panel C

93

0

10

20

30

40

50

60

70

80

90

0 0.05 0.1 0.15 0.2

Lateral opening of shiplap joint, in

Load

, kip

sK = 1367 k/in

Stiffness change at 22 k

Figure 5.22 Lateral opening of center shiplap joint for the inelastic test of Panel C

Lateral opening of the center shiplap joint was monitored during the test and Figure

5.22 exhibits the load versus lateral deflection. Initial displacement, albeit small, occurred at

3 kips. Since the conditioning load was carried to an errant peak of 30 kip (6 kip above the

target load), the possibility may exist that the concrete cracked longitudinally above the

shiplap joint before the inelastic test. The initial stiffness of 1367 kips per inch was

maintained until 22 kips, where a significant change in stiffness arises. After 35 kips, the

stiffness remained relatively constant until the estimated yield point of 80 kips. The LVDT

reached it’s measuring limit at about 77 kips.

The same structural behavior seen in the previous concrete deck panels such as

cracking patterns and FRP panel deflection, combined with the observation that capacity was

independent of span (96, 86, and 91 kips for Panels A, B, and C, respectively) provided

enough evidence to conclude a punching shear failure. Figure 5.23 is a post mortem concrete

cut through the center span of Panel C and it strongly supports a punching shear failure

94

mode. This cut section exposed the internal concrete cracking pattern that evolved directly

from the edge of the loading patch. The cracking pattern embodies the typical 45° angle of

the principal tension expected in a punching shear failure. Notice how the punching shear

cracks start, or end, at the shiplap joints of the FRP deck forms.

Figure 5.23 Post-Failure Panel C Showing Punching Shear Failure Under Load Patch

5.3 Positive Moment Beam Test

The positive moment test was designed to isolate the positive moment characteristics

in the primary deck direction, such as strength per foot of width, of the FRP reinforcement

system. In order to do this, each specimen was subjected to a line load, at mid-span, to

complete a 3-point bending test. Ultimately, the experimental results were used as a

benchmark needed to help build a finite element model of the FRP system to find the

Edge of Load Patch

Top of Corrugation

Shiplap Joint

Denotes top of cut

95

continuous slab moment distribution behavior that is difficult to determine through

laboratory testing.

To reiterate, each of the three positive moment beams, all with an 8’-8” span, was

sequentially subjected to an elastic and inelastic testing regime. Running the elastic test,

described in Section 4.3, naturally conditions the beam to a “serviced” state. The elastic

steps required digital data recording, however, the data was only used to check the

instruments. In addition, there were no significant visual observations because of the low

level of loading and therefore the elastic testing was straightforward. Hence, unless

noteworthy, test observations and information on the elastic testing step not covered in

Chapter 4, will not be further discussed here.

The “lion’s share” of data analysis, and therefore testing significance, falls upon the

inelastic testing step. Each positive moment beam was loaded monotonically up through and

past the ultimate capacity until a significant deflection was attained, to ascertain a post-

ultimate behavior. During the course of loading, new and growing concrete surface cracks

were marked with their particular load level. Audio cracking noises were also noted during

testing. Once again, the LabView software was utilized to display real-time load versus mid-

span deflection information for the operator to monitor as the test was conducted.

5.3.1 Beam D1

One piece of information from the elastic test should be noted, the first cracking

sound occurred at 18 kips of the first cycle the beam experienced. For the inelastic test, a

manual strain gauge apparatus, location described in Section 4.3, was used to verify the

measured electrical strains. The manual strain gauge information was recorded at 0 kips, 24

96

kips and at a deflection of 1.67” and recorded in the lab book. No other information was

recorded.

0

10

20

30

40

50

60

70

0 0.5 1 1.5 2 2.5 3 3.5 4

Stroke deflection, in

Load

, Kip

s

P yield 54.4 kips

K effective = 118 k/in

Figure 5.24 Load versus actuator stroke for the inelastic test of Beam D1

At about 34 kips a sharp crack was heard, but unseen; it appeared to originate from

the bottom surface of the beam. Yield was estimated at 54.4 kips with an initial stroke

stiffness of 118 kips per inch as seen in Figure 5.24. Ultimate capacity occurred at 63.5 kips

and at stroke deflection of 0.90”. As for the post-ultimate behavior, the beam was loaded for

a very significant deflection beyond the ultimate load. At a deflection of 2.40” and carrying

56 kips, one end of the beam rotated enough to begin to bear upon the front edge of the

support block, effectively changing the span length. Examining the saw tooth pattern seen

above the deflection of 2.5”, in Figure 5.24, was apparently a result in strength degradation.

At a deflection of 3.25”, the FRP deck forms separated from the concrete above them,

essentially delaminating. The test was carried out to a final deflection of 3.72” and carrying

a load of 47 kips. The test was terminated because of safety concerns. Although not plotted

97

in this report, it should be noted that stroke and potentiometer deflections (4” toward the

support from the centerline of the span) were plotted together and very little discrepancy

between the two readings was found.

3

3.2

3.4

3.6

3.8

4

-0.0015 -0.001 -0.0005 0 0.0005 0.001 0.0015

0

1

2

3

4

5

6

7

8

-0.008 -0.006 -0.004 -0.002 0 0.002 0.004 0.006 0.008 0.01

Strain, in/in

Dep

th o

f Sec

tio

n,in

20 kip 30 kip 40 kip 50 kip 60 kip 63 kip

Concrete,Strain Gauge 2

FRP,Strain Gauge 4

Blow up of neutral axis

Figure 5.25 Depth of section versus material strain as a function of load, Beam D1

Material strains as a function of section depth for Strain Gauge 2 on the top surface

and Strain Gauge 4 on the bottom surface are graphed in Figure 5.25. If plane sections are

assumed, the plot clearly illustrates the neutral axis maintaining its position at around 3.4”

down from the top surface throughout the test. Surprisingly, at beam failure the concrete

strains reach twice the unconfined compression strain, 0.003, accepted for design by

American Concrete Institute. Concrete is not a true linear elastic material, however, so the

98

strains do not translate into stress at a linear rate, especially near failure (due to flexural

concrete crushing). The FRP material strains approach 0.01 at failure. The other set of set of

strain gauges (Strain Gauge 3 and Strain Gauge 5), located at the same distance from the

centerline, produced strains very similar in magnitude. Although, similar values from strain

gauges at the same distance from center does not disprove the possibility of systematic

errors. Using the uniaxial linearly elastic stress-strain relationship, σ = ε*E, and modulus of

elasticity’s of 4840 ksi for the FRP (see Table 4.1), the bending stresses at failure would be

around 43.5 ksi (neglecting the contribution of selfweight). As previously stated, it was a bit

more difficult to translate concrete strains into stress near failure. To verify these large strain

values for the concrete, a mechanical strain gauge, complementary to the location of Strain

Gauge 3, was used. The mechanical strain gauge measured a concrete strain of 0.0026 and

the electrical strain gauge read 0.0016 at 24 kips. There seems to be some discrepancy

between readings of the two instruments, however, given the accuracy of the mechanical

strain gauge it was hard to derive a definitive conclusion. It was unfortunate not to have

more data points in the elastic region for comparison. In Beam D3 test, many more

mechanical strain gauge data points were taken to provide a better data set for independent

verification.

Failure was due to top concrete crushing due to flexure near mid-span, as the section

was designed to perform. Figure 5.26 shows the mid-span region on each sides of the beam

where the concrete crushing failure occurred. The vertical concrete cracking patterns, caused

by bending, on each side of the beam look quite similar. The excessive deflection, beyond

the 0.90” ultimate deflection, caused the large horizontal crack seen on the Picture A of

Figure 5.26. As explained in Chapter 1, aggregate was epoxied to some, but not all

99

horizontal areas on the top surface of the FRP deck form, see Figure 1.2, in order to transfer

horizontal shear between the FRP deck form and the concrete. The Picture A shows the

flexure cracking pattern above the side of the beam that has the aggregate epoxy coated

outboard side of the FRP deck form and the Picture B did not have the aggregated epoxy

coated attached to the outboard of the FRP deck form.

A B

Figure 5.26 The concrete crushing near the load for both sides of Beam D1

5.3.2 Beam D2

During the elastic test, the first cyclic load to 24 kips produced a concrete bending

crack at about 21 kips, near the centerline, on the outboard side of the beam with no bonding

(area that lacks aggregate epoxied to the top surface of the FRP deck form) between the FRP

deck form and the concrete. The opposite side of the beam, the outboard side with bonding

between the FRP deck form and the concrete, experienced no cracking during the elastic

testing.

100

0

10

20

30

40

50

60

70

80

90

0 1 2 3 4 5Stroke displacement, in

Lo

ad, k

ips

P yield = 66 kip

K eff = 102 k/in


This time around, detailed visual and audio information was kept for the inelastic test.

Not much occurred in the elastic region of the inelastic test. Figure 5.27 is the graphic

representation of the load versus stoke relationship. At just around a load of 54 kips a loud

“snap” was heard from below the beam, and in hindsight, this marked the very beginning of

the ultimate “region”. A second loud “snap”, again, originated from the underside of the

beam at about 64 kips. At 65 kips, 0.68” deflection near mid-span, a large horizontal shear

crack developed on one side of the beam (see Figure 5.29 A). Initial effective stress was

estimated at 102 kips per inch and the estimated yield point was calculated as 66 kips.

Ultimate capacity was reached at a load of 77.6 kips (23% higher than D1) and a deflection

of 1.02” under a flexure concrete crushing mode. The test continued on into the post-

ultimate region. At 1.45” deflection, the concrete under the 3” square bar crushed

completely. At 1.75”, the side of the beam with the non-bonded outboard side of the FRP

deck form started to delaminate. At 2.77”, another loud snap was created from the FRP deck

101

forms, which may have been starting to de-bond fully from the concrete. At 3”, still carrying

50 kips, the FRP deck form (from the side of the beam with the non-bonded outboard side)

became fully de-bonded, and near mid-span, a large ½” wide flexural crack developed. The

inelastic test was finally terminated at a deflection of 4” still carrying a load of 50 kips.

0

1

23

45

6

78

-0.012 -0.008 -0.004 0 0.004 0.008 0.012Strain, in/in

Dep

th o

f Sec

tion,

in

20 kip 30 kip 40 kip 50 kip60 kip 70 kip 76 kip (Ult.)

Concrete, StrainGauge 2

FRP, StrainGauge 4


Material strains as a function of section depth for the same strain gauges used for

Beam D1 are graphed in Figure 5.28. If plane sections were assumed, the plot clearly

illustrates the neutral axis migrating down as the load increased. Again the magnitude of the

concrete strain at beam failure was very high, 0.011. The concrete strains are over four times

the unconfined compression strain accepted for design by American Concrete Institute. A

mechanical strain gauge was not used in this test to measure concrete strains. Therefore, it

was difficult to independently verify the high concrete strain readings and disprove any

systematic errors. The high concrete strains may be due to the top bi-directional grid

102

confining the concrete. This possibility may be unlikely, because under the same failure

mode, Bank et al (1992) found concrete strains no greater than 0.006 at ultimate.

The FRP material strains approached 0.012 at beam failure. As before, the other set

of strain gauges located at the same distance from the centerline produced very similar

values. Again, using the linearly elastic stress-strain relationship and modulus of elasticity of

4840 ksi for the FRP, the stress of the FRP deck form at beam failure was be 54.5 ksi

(neglecting the contribution of selfweight).

The cause of failure of this beam was a bit more difficult to diagnose. It seems to

have failed from a combination flexure and shear failure on the side of the beam with the

bonded outboard side of the FRP deck form, see Picture A in Figure 5.29. This side of the

beam also developed a series of well-spaced flexural cracks. As the deflections increased,

the tops of the flexural cracks were joined by a horizontal shear crack that formed just below

the top FRP grid panel. Picture B of Figure 5.29 shows the non-bonded outboard side of the

beam with a few large flexure cracks, which are indicative of a lack of horizontal shear

transfer between the FRP deck form and the concrete.

103

Shear Crack

A B

Figure 5.29 Flexure-shear cracking on left and flexure cracks on the right, for Beam D2

Something happened in testing Beam D2 that did not occur in testing of Beam D1.

Figure 5.30 shows two chunks of concrete missing from one end of the beam. Near the end

of the test, well into the post-ultimate region, two cone shaped chunks of concrete actually

broke off and landed many feet away from the specimen. When they dislodged from the end

of the beam they also made a loud “boom”, as though a small caliber gun discharged.

Looking closely, the webs of the corrugations were split longitudinally separating the

corrugation into top and bottom pieces. At this stage in the test, post-ultimate, the deck

forms were completely de-bonded from the concrete and, therefore, carrying the load alone.

Through bending, this would have forced large horizontal shear forces in the web of the

corrugations and under the large deflection, the web sheared longitudinally. When the webs

sheared, they forced the concrete on the ends to explode off the supported end of the beam.

104

Split in the corrugations

Figure 5.30 Two chunks of concrete dislodged at the end of Beam D2

5.3.3 Beam D3

Unlike the previous two positive moment beam test, there were no significant events

noted during the elastic (conditioning) test. The high concrete strain values measured in the

previous tests induced the addition of a series of mechanical strain gauge readings to

independently verify the strains. Another difference from the previous beam tests was the

addition of two strain gauges, one top (Strain Gauge 6) and one bottom (Strain Gauge 7), 12

inches from the center of the span and centered in beam width. This was another attempt to

provide independent strains for comparison to the anticipated high strains produced near the

center span. The strain gauge locations for Beam D3 are illustrated in Appendix C. One

additional difference between Beam D3 and the previous two positive moment beams was

the lack of a top bi-directional grid. The bi-directional FRP grid was excluded in the

105

reinforcement of this beam to explore its impact on performance in the positive moment

region.

0

10

20

30

40

50

60

70

80

0 0.5 1 1.5 2 2.5

Stroke Deflection, in

Load

, kip

s

P yield = 68 kip

K eff = 126 kip/in


The inelastic beam test was conducted in the same manner as the previous positive

moment beam tests. Figure 5.31 exhibits the load versus stroke relationship for Beam D3.

Strangely, the first crack heard occurred at around 16 kips. One would think that the elastic

testing would have “seated” the beam from cracking in this region in the inelastic test. The

yield point was calculated at about 68 kips. The initial effective stiffness was 126 kips per

inch, which was a bit higher than what was produced by the previous two positive moment

beams. A large flexure/shear crack developed on the un-bonded outboard side of the beam at

about 70 kips (see Figure 5.35B). These flexure/shear cracks significantly expanded in size

at a deflection of 0.88”, which was right at the ultimate load of 76 kips. The load was

applied through 2” of stroke deflection holding 33 kip. The post-ultimate capacity of this

beam was considerably below the previous two positive moment beams. This may be

directly attributed to the lack of compression capacity, since the top bi-directional FRP grid

106

was missing. Comparing cracks in Figure 5.29A (Beam D2) and the cracks for Figure 5.35B

(Beam D3) leads to the belief that the horizontal cracks seen in Figure 5.29A, which are not

seen in Figure 5.35B, were related to the top grid. It was possible that after the FRP panel

de-bonded from the concrete, the capacity of Beam D3 (without top grid) dropped because

the FRP deck forms alone provided resistance. In Beam D1 and D2, the top grid provided an

additional resisting entity, through tension reinforcement for the top surface of concrete, in

addition to the FRP deck forms.

012345678

-0.012 -0.008 -0.004 0 0.004 0.008 0.012Strain, in/in

Dep

th o

f sec

tion,

in

20 kip 30 kip 40 kip 50 kip 60 kip 70 kip 76 kip (Ult.)


FRP,Strain Gauge 4


Graphed in Figure 5.32 were the material strains through the depth of section, as a

function of load for Strain Gauges 2 and 4 (SG4 was below the outboard side of FRP deck

form). These values were measured 4” from the center of the span. Assuming plane

sections, the neutral axis location remained almost constant with a slight drop in position as

the load reached ultimate. Again, the material strains, especially the concrete, were very

high. Fortunately, there were additional sources to compare strain values. First, the other

107

pair of Strain Gauges, 3 and 5, that were positioned at the same distance away from the

center span as the gauges graphed above. These values produced by Strain Gauges 3 and 5

were similar to the strains in Figure 5.32. Since the high concrete strains seen in Beam D3

were comparable to the concrete strains seen for Beams D1 and D2, the top grid did not

create the high strains by confining the concrete.

01020304050607080

-0.012 -0.01 -0.008 -0.006 -0.004 -0.002 0Concrete Strain, in/in

Lo

ad, k

ips

Strain Gauge 2 Strain Gauge 3 Mechanical

Figure 5.33 Relationship of 3 concrete strain gauges at the same distance from the centerline

The mechanical strain gauge attached to the top concrete surface was complimentary

in position to Strain Gauge 3. Figure 5.33 shows the close relationship of these sources of

concrete strains, particularly as the load passed 30 kip. The mechanical strain gauge results

may suggest that concrete strain values above 0.0055 may not be accurate and values above

this may be false readings.

As seen in Figure 5.32, the strains in the FRP deck form at beam failure were 0.01,

which (uniaxial linearly elastic stress-strain relationship, σ = ε*E, and modulus of elasticity’s

of 4840 ksi for the FRP) when neglecting stress due to selfweight, translates into a stress of

48.4 ksi.

108

0

1

2

3

4

5

6

7

8-0.012 -0.008 -0.004 0 0.004 0.008 0.012

Strain, in/in

Dep

th o

f se

ctio

n, i

n

20 kip 30 kip 40 kip 50 kip 60 kip 70 kip 76 kip (Ult.)


FRP,Strain Gauge 7

Figure 5.34 Material strains through the depth of section for 12” away from the centerline

Another comparison of material strains was from the pair of strain gauges, SG6 and

SG7, 12” away from the centerline. At this location, the strain values through the depth of

section as a function of load are graphed in Figure 5.34. As opposed to the behavior of the

neutral axis at the center span, seen in Figure 5.32, the neutral axis traveled up as the load

increased, if plane section behavior is assumed. This set of strain gauges produced a

maximum concrete strain of 0.004 at beam failure, which seems a bit more reasonable (this

concrete strain value was lower because it was measured 8” further away from the center

span than the values seen in Figure 5.32). Taking into account the mechanical strain gauge

readings and the reasonable readings from SG6, the concrete strains above 0.006 were not

accurate for all the Beams D.

109

A B

Large Flexure/Shear Crack

Figure 5.35 Horizontal shear and flexural-shear cracking of Beam D3

The mode of failure for this beam can be directly linked to the absent bi-directional

FRP grid. The large horizontal shear crack in Picture A and the large flexural-shear crack in

Picture B of Figure 5.35 may not have developed, and certainly would not be as wide, if the

FRP grid was present. This mode of failure was not seen in the previous two positive

moment beams, consequently, bolstering an argument for the importance of the FRP grid in

the positive moment region.

Another observation taken from the test was the difference of concrete cracking

patterns seen on each side of the beam. Picture A shows a large horizontal shear crack that

developed along the beam at the height of the corrugation of the FRP deck form. This crack

would of developed into a flexural/shear crack if this side of the beam was bonded between

the concrete and the FRP deck form outboard. On the side of the beam with the epoxy-

coated FRP deck form, Picture B of Figure 5.35, many flexural cracks developed at mid-

span. The other side, the side without epoxy-coated aggregate, had very few flexural cracks.

This lack of flexural cracking on the side of the beam was indicative of a lack of horizontal

shear transfer between the two materials. Following this line of reasoning, the beam must

110

have been acting as partially composite, thus reducing the strength and efficiency of the

section.

5.4 Negative Moment Beam Test

The main goal of the negative moment beam test was to isolate the bending

characteristics provided by the top FRP grid reinforcement. Specific characteristic of interest

include the strength per foot of width. In order to do this, each specimen was subjected to a

line load, at each mid-span, of the 2-span continuous beam. The experimental results were

used to help build a finite element model of a prototype bridge deck system. The goal of the

FEM model was to find a negative moment distribution width, which is difficult to determine

through laboratory testing.

Each of the three negative moment beams had two identical 8’-0” spans. Each beam

was subjected to inelastic testing only. Each span of the negative moment beam was loaded

monotonically and simultaneously up through and past the ultimate capacity until a

significant deflection was attained, thus ascertaining a post-ultimate behavior. During the

course of loading, new and growing concrete surface cracks were marked. Audio cracking

noises were noted during testing. Once again, the LabView software was utilized to display

real-time load versus mid-span deflection information for the operator to monitor as the test

was conducted. The method used to conduct the inelastic testing steps was described in more

detail in Section 4.4.

5.4.1 Beam E1

As noted in the description of test method, it was quite difficult to place the negative

moment specimens into the test assembly. In placing Beam E1, it was fortuitous that the

111

beam was placed without developing any concrete cracks in the negative moment region.

Unfortunately, this was not the case for all of the negative moment beams. Anticipating

negative moment concrete cracking over the middle support early in the test, the width of the

first concrete crack was measured by a magnified scale at set load increments. The loads

applied to and the deflections of each span were digitally recorded for later data analysis. In

addition, to monitor the strains in the negative moment region, three strain gauges were

placed through the section of the beam at the center of the middle support. One strain gauge

was place on the concrete surface, another attached to the bi-directional FRP grid, and one

gauged the concrete bottom of the section over the middle support (see Figure 4.22 for a

detailed cross-sectional view). The FRP deck form was not continuous over the support.

0

10

20

30

40

50

60

70

80

0 0.2 0.4 0.6 0.8 1 1.2 1.4 1.6

Center span LVDT deflection, in

Lo

ad, k

ips

Span A Span B

P yeild = 66

K eff = 249 k/in

Figure 5.36 Load versus deflection curves for Span A and Span B for Beam E1

For the load versus LVDT deflection curve seen in Figure 5.36, the 200-kip actuator

loaded Span A and the 55-kip actuator loaded Span B of the continuous beam. An unknown

error in the LabView software delayed the recording of instrument data until just after 8 kip

had already been applied, therefore, the initial readings at no load were not recorded. This

112

introduced an unknown level of inaccuracy for response during the first 8 kips of load within

the presented data for Beam E1. The LVDTs were offset close to zero at the beginning of the

test but the stoke readings were not. All measurements relative to the condition at 8 kips are

exact.

Test notes indicate that the first negative moment concrete crack on the top of

surface, over the middle of the support, occurred at a nominal load of 7.8 kips and for the rest

of the test this crack width was measured at various load intervals. Unfortunately, due to the

initial recording error, the loss in stiffness at 7.8 kips cannot be graphically identified in the

load versus deflection relationship. A second negative moment crack developed at 17 kips,

just a few inches away from the first the support center (see Figure 5.38), but this crack width

was not monitored. Both spans were loaded simultaneously up to 50 kips. At 50 kips, Span

B (the 55 kip actuator) reached its limit of possible load application. Therefore, the 50 kip

load was maintained in Span B while Span A continued to be loaded up to failure. Failure

occurred abruptly at a load of 73.3 kip in Span A and a load of 50 kip in Span B. As done on

earlier test specimens, the standard method (Preistley, 1992) calculated an effective stiffness

of 249 kips per inch with a yield load of 66 kips for Span A.

113

-0.016-0.014-0.012-0.01

-0.008-0.006-0.004-0.002

00.0020.0040.006

0 10 20 30 40 50 60 70 80

Load of 200-kip actuator, kip

Str

ain

, in

/in

Top FRP grid Bottom

Compression

Tension17 kips

Figure 5.37 Strains from the section centered over the middle support Beam E1

The strains measured through the depth of a section centered over the middle support

are graphed in Figure 5.37. The first and most notable observation was the fact that the top

surface concrete strain gauge did not register any significant strains. As seen in Figure 5.38,

the top surface strain gauge was place right near the negative moment concrete crack over the

middle support. This concrete crack likely prevented significant tension strains from

developing. Once the crack developed, the concrete directly adjacent to the crack was

relieved of strain and, therefore, did not carry tension stress.

114

1st crack 2nd crack

Center of middle support

Figure 5.38 Top surface strain gauge between the negative moment cracks, Beam E1

Looking closely at Figure 5.37, the abrupt increase in the rate of strain for the FRP

grid at 17 kip load in each span leads one to conclude that this was the point (just when the

second concrete crack appeared) that all tension resistance was transferred from the concrete

to the FRP grid. It is interesting to note that rate of strain slowly dropped off after the abrupt

increase at 17 kip. The grid reached a tension strain of 0.0134 at the time of ultimate

capacity. Using a linear uniaxial stress-strain relationship for the FRP grid, σ = ε*E, the grid

stress at the ultimate capacity of the beam (E = 4480 ksi) reached 60 ksi. The FRP grid stress

at failure was well below the tensile strength of 83 ksi. The concrete strains on the bottom of

the section show a nice, somewhat linear, compression strain profile all the way up to

ultimate load in the beam. The compression strain on the bottom of the section over the

middle support at beam load capacity reached 0.0036.

115

0

1

2

3

4

5

6

7

8-0.008 -0.004 0 0.004 0.008 0.012

Strain, in/in

Dep

th o

f sec

tion,

in


Tension in FRP grid

Compressionin Concrete

Figure 5.39 Strains through the depth of section as a function of load for Beam E1

The stress through the depth of section as a function of the load provided by the 200-

kip actuator is illustrated in Figure 5.39 (assuming plane sections). Until the second negative

moment concrete crack developed, the neutral axis was very close to the centroid of the

section. Right after the second crack, the neutral axis dropped down dramatically. As the

load increased toward ultimate, the neural axis slowly moved up to a little over an inch above

the top of the bearing.

Table 5.1 Measured widths of the first concrete crack to develop on the top surface

Load, kip Crack Width, in

7.8 0.005 12 0.01 16 0.015 22 0.020 30 0.03 40 0.03 50 0.04

116

Table 5.1 lists the growth measurements of the first negative moment concrete crack

(see Figure 5.38) widths in Beam E1 at various load intervals. The rate of crack growth

seems to be consistent with the rate of load.

Debonding

Note: 3” load bar was here

A B Debonding

Figure 5.40 Failure of Beam E1 for each side of beam in Span A.

Figure 5.40A shows a massive shear failure between mid-span and the middle

support, which basically ripped through the bi-directional FRP grid in Span A. This shear

failure was very similar to failure reported by Bank et al. (1992) on a double span specimen.

The failure was very quick and there were no clear indications that failure was imminent.

Unfortunately, the Picture A was slightly out of focus to see that the FRP deck form actually

de-bonded from the concrete at the middle support. Picture B clearly shows the shear crack

that developed near the point of load application on the side of the beam without bond

between the concrete and the FRP deck form. In Picture B, the shear crack developed from

the point of load, in the center of Span A, and worked toward the FRP deck form. Once the

shear crack reached the deck form, delamination occurred between the deck form and the

concrete all the way to the middle support. The shear crack on this side of the beam has the

bonded epoxy-coated aggregate outboard side of the FRP deck form. De-bonding of the FRP

deck forms occurred just after shear failed the beam. What is quite noteworthy was the fact

117

that the failure occurred somewhere other than the negative moment region. The weakness

of the beam under these test conditions was not the grid, and ultimately, the negative moment

capacity was not reached and remains unknown.

5.4.2 Beam E3

All of the previous testing procedures, such as measuring the widths of the first

negative moment concrete crack over the middle support, were applied to the test of Beam

E3. Again, with caution, the beam was placed into the test position without causing a

concrete crack in the negative moment region of the beam. Therefore, like Bean E1, at the

start of the test the beam was initially uncracked. Strain gauges were in the same position

through the depth of section over the middle support as seen for Beam E1.

0

10

20

30

40

50

60

70

80

0 0.5 1 1.5 2 2.5

Center span LVDT deflection, in

Load

, kip

s

Span A Span B

P yield = 61 k

K eff = 272 k/in

Figure 5.41 Load versus deflection curves for Span A and Span B for Beam E3

Figure 5.41 exhibits the load versus deflection curve for both spans. Figure 5.42

shows the first negative moment concrete crack over the middle support that occurred at 12

kips and the second concrete crack arose at 22.5 kips. Loading was applied simultaneously

118

to 50 kips, when Span B held a constant 50 kip load and Span A continued loading until

beam failure. Failure of the beam occurred in Span A at a load of 67.5 kips. The beam

failure was not abrupt, as seen in Beam E1, but more drawn out, holding a considerably large

load through noteworthy deflections. Between the deflections of 0.75” and 1” (post-ultimate

region), it was seen that the bond between the FRP deck form and the concrete was lost near

the middle support in Span A. Figure 5.41 doesn’t reveal all the deflection data because the

data recording was stopped at 3.2”, however, the test was carried out until a final deflection

of 5”. The test was terminated out of safety concerns. From the data and using the Priestley

method, the yield load for Span A was calculated at 61 kips with an effective stiffness of 272

kips per inch. The values were similar to measurements taken for Beam E1.

1st crack

2nd crack

Center of middle support

Figure 5.42 The first two concrete cracks over the middle support, Beam E3

119

-0.014

-0.012

-0.01

-0.008

-0.006

-0.004

-0.002

0

0.002

0.004

0.006

0.008

0 10 20 30 40 50 60 70 80

Load, kip

Str

ain

, in

/in

Top FRP grid Bottom

Compression

Tension1st crack

2nd crack

Figure 5.43 Strains from the section centered over the middle support, Beam E3

Once again, the top strain gauge on the surface measured little concrete tension strain.

The strain gauge was near the tension cracks over the middle support and once those cracks

occurred, the concrete tension strains were relieved and remain small. Evaluating the bi-

directional grid strains in Figure 5.43, there were noticeable increases in the rate of

stress/strain change at two points during the test. Both points (or levels of load) occurred at

the observed negative cracking loads over the middle support, at 12 and 22.5 kip. Like Beam

E1, this means that at the two concrete cracking episodes, the tension strength shifted to the

bi-directional grid (although, the concrete was not expected to carry the load in tension). The

strain in the grid at ultimate load of the beam was 0.0119 and this strain was translated into a

stress of 53 ksi (σ = ε*E), which was below the result of Beam E1.

The concrete strains on the bottom of the section show a nice, somewhat linear,

compression strain profile all the way up to failure of the beam. The concrete compression

strain on the bottom of the section over the middle support at beam failure reached 0.0053.

120

0

1

2

3

4

5

6

7

8-0.008 -0.004 0 0.004 0.008 0.012

Strain, in/in

Dep

th o

f se

ctio

n, i

n


Tension in FRP grid

Compressionin Concrete

Figure 5.44 Strains through the depth of section as a function of load, Beam E3

Figure 5.44 shows the strains of the material through the depth of the section over the

middle support, assuming plane sections, as a function of the load in Span A. Once the

concrete cracked in the negative moment region, the neutral axis dropped from about the

centroid down to 1.4” up from the bottom of the beam. Then as the load increased in Span

A, the neutral axis rose slightly.

Table 5.2 Measured widths of the first concrete crack to develop on the top surface, Beam E3

Load, kip Crack Width, in

12 0.01 16 0.015 20 0.02 30 0.02 40 0.03 50 0.045

Table 5.2 lists the growth measurements of the first negative moment concrete crack

width in Beam E3 at various load intervals. The rate of crack growth seems to be consistent

with the increase in load.

121

A Three Shear Cracks

B

Figure 5.45 Each side of Beam E3 showing a shear failure occurring at 67 kips, Beam E3

Figure 5.45 reveals three major shear cracks in Span A that failed Beam E3 at a load

of 67 kips. The major shear cracks in Picture A developed at or very close to the same

moment. Once the beam failed in shear, the FRP deck form de-bonded from the concrete

near the support. Picture A was the side of the beam that had the bonded epoxy-coated

aggregate outboard side of the FRP deck form. Picture B shows one major flexure crack near

mid-span (which forms a hinge before the shear failure) and this was the side of the beam

that had the non-bonded, or smooth, outboard side of the FRP deck form. From these

pictures and the failure pictures of Beam E1, one could conclude that the type of shear

cracking that developed was dependent on whether this bond existed.

Another main point that can be drawn from the two negative moment beam tested

was that the beams did not reach the negative moment capacity.

5.4.3 Beam E5

The test of E5 had unknown problems with the data acquisition process and all the

data was unusable. While being place into the test position, beam E5 incurred a large

negative moment concrete crack in the region where the beam would rest on the middle

122

support. There were some test notes that provide useful information from the test. These

notes were now very valuable in the absence of any useful digital data.

Since there was an initial concrete crack over the middle support, no new cracks were

observed in this region at relatively low loading, although, the initial cracks did grow. At 30

kips, we momentarily paused testing to change a voltage divider to try and remedy the data

recording errors. Nevertheless, this did not fix the problem. The test continued on and at

about 48 kips the concrete began to crush, in bearing, under the 3”-square loading bar in

Span A. At 50 kips, the 55 kip actuator held this load in Span B, while the 200 kip actuator

continued to load Span A. At 54 kips, a positive moment flexure crack developed, and thus a

hinge, at mid-span on the side of the beam in Span A, see Picture B in Figure 5.46 (this was

the side of the beam without aggregate epoxied to the outboard side of the FRP deck form).

Shear failure occurred at a load of 72 kips in Span A, see Picture A in Figure 5.46.

A B

Large Flexurecrack

Dowel Crack

Figure 5.46 Concrete cracks on each side of Span A of Beam E5

Figure 5.46 reveals the concrete cracking on each side of the beam. Picture A is a

shear crack that caused the failure, which developed on the side of the beam with the

aggregate epoxy coated to the outboard side of the FRP deck form. Picture B shows the large

123

flexure crack that developed before failure on the side of the beam without aggregate epoxied

to the outboard side of the FRP deck form. Just after the shear failure in Picture A, a large

horizontal crack developed by pseudo dowel action, which means the grid carried the post-

failure shear. This horizontal crack traveled from the point of load to the middle support.

Both cracks were almost identical to the failure cracking seen on the two previous negative

moment beams. It has become clear that the presence or absence of epoxy coated aggregate

on the FRP deck form influences the cracking behavior at failure.

A

Concrete removed in Picture B

B

Figure 5.47 Cracking at failure on side of beam without aggregate epoxied to FRP form

The pictures of Figure 5.47 show the side of the beam without aggregate epoxy

coated to the outboard side of the FRP deck form in Span A. After the test was complete,

large chunks of concrete were removed by hand, thus exposing the smooth outboard surface

and the loss of bond between concrete and the FRP deck form.

124

A

Shear Crack

B

Figure 5.48 Additional cracks of Beam E5

Figure 5.48 provides some additional views of the concrete cracking that occurred in

Beam E5. Picture A is a close up of the shear crack from the bearing point that traveled

toward the loading point (on the side of the beam with aggregate epoxied to the outboard side

of the FRP deck form). Picture B is an elevated view of the concrete cracking on the top

surface of the beam in the negative moment region over the support.

5.5 Additional Concrete Deck Panel Test, Panel C2

After running through the regime of testing described so far, it was concluded that the

FRP deck form was inefficient in composite action because of poor bond on surfaces lacking

the epoxied aggregate. Increasing the coverage of aggregate to all horizontal surfaces of the

FRP deck form should boost performance. To test this hypothesis, an additional concrete

deck panel was constructed, but this time the FRP deck form was covered with aggregate on

all horizontal surfaces before casting the concrete. The increase of aggregate coverage

should increase the bond and horizontal shear capacity and in turn increase panel strength.

The additional concrete deck panel, Panel C2 (same geometry as Panel C), was also chosen

to test the inclusion of a 4” wide x 1” deep FRP Fibergrate Molded Grating (see Figure

125

3.10) over the shiplap joint to provide additional strength against the lateral opening of the

shiplap joint under load. Built-in strain gauges through the depth of section at the center of

span near the load point were also added. There were 5 strain gauges through the 8” depth,

thus providing a detailed material strain profile of the section. In Section 4.6, there is a

description of the exact locations of the strain gauges within the section and position relative

to the loading patch.

Panel C2 was subjected to the same testing procedures as conducted on Panel C.

There was a different scheme for instrument location, however, especially for the strain

gauges. The shiplap joint in the center was again monitored for lateral opening. There were

no noteworthy issues that arose during the conditioning or during the elastic testing steps.

0.0

20.0

40.0

60.0

80.0

100.0

120.0

140.0

0.0 0.5 1.0 1.5 2.0 2.5 3.0 3.5Stroke deflection, in

Load

, kip

s

P yield = 106 kip

K effective = 217 kip/in

Figure 5.49 Load versus actuator stroke for the inelastic test of Panel C2

Figure 5.49 shows the stroke versus the load profile. The stroke deflection at 20 kips

was 0.136”, which equals L/706. The difference between the stroke and the potentiometer

deflection measurements, seen in Figure 5.50, reveal a nice linear disparity up to 120 kips.

126

The yield point was calculated at 106 kips, a 33% increase over Panel C. The initial stroke

stiffness was calculated at 217 kip per inch, which was a 43% increase over Panel C. At

about 120 kips the reaction frame of the 200 kip actuator began to lift into the air, because

one base plate started to yield. Substantial, and continuous, cracking and popping noises

started at 120 kip. At a load of 130 kips, it was noticed that the frame was lifting and the test

was stopped out of safety concerns. The resistance capacity had not dropped at the end of the

test, although the uplift of the reaction frame prevented further loading. The maximum

reliable stroke deflection (reaction frame began to lift off the structural floor) reached 0.82”

at 120 kips. For reasons stated earlier, there was not post-ultimate testing.

0.0

20.0

40.0

60.0

80.0

100.0

120.0

140.0

0.0 0.5 1.0 1.5 2.0

Difference between stroke and potentiometer, in

Load

, kip

s

Figure 5.50 Difference between the stroke and potentiometer readings for Panel C2

127

Punching Pocket

Approx. Location of SIP Shiplap Joints

Figure 5.51 Top surface cracking pattern of Panel C2

As seen in Figure 5.51 the surface concrete cracking pattern was much different then

the oval cracking patterns documented for the original three concrete deck tests. Punching

shear failure was achieved since there was an indented punching pocket under the load patch.

Like the previous tests, there was no concrete cracking along the unsupported span edges of

the panel, leading to the conclusion that there was little load distributed to the edges. Due to

shear forces the supported ends of the panel did see a major increase in large vertical

concrete cracking. In general, the longitudinal cracking pattern on the surface, especially the

cracks on the left hand side of Figure 5.51, was above the position of the shiplap joints

between the FRP deck forms below. As in the previous concrete panel tests, the two middle

128

FRP deck forms deflected more than the adjacent set of two forms to either side. Figure 5.52

exposes the degree of differential deflection between these forms.

Differential Deflection Between Deck Forms

Figure 5.52 Differential deflection at the shiplap joints of the center FRP deck

forms in Panel C2

129

-2.50

-2.00

-1.50

-1.00

-0.50

0.00-54.0 -42.0 -30.0 -18.0 -6.0 6.0 18.0 30.0 42.0 54.0


Def

lect

ion

, in

.

16 kip 40kip 60 kip 80 kip 100 kip 120 kip 130 kip

CenterEdge Edg

Figure 5.53 Deflection profile of Panel C2 along the center of span

Figure 5.53 represents the deflection of the center of the span at various load

increments. The plotted data was from LVDTs, potentiometer (9” off center) and the stroke

(center). The reaction frame lifted off the structural floor, and in the process lifted the

actuator with it and this has, in part, inflated the center deflection near ultimate. Figure 5.51

showed the concrete crushed directly under the load patch; this also contributed to the

significant difference in displacement values of the stroke and potentiometer values at

ultimate. This was just the opposite of what has happened with the corners lifting off the

reaction blocks in the original three concrete deck panels. The test of Panel C2 did not see

the four corners lifting of the supports.

As mentioned earlier, Panel C2 used a different strain gauge profile scheme than what

was chosen for the original three deck panels. Instead of using a series of two strain gauges

(one top and one bottom) positioned along the center of span, there were five strain gauges

130

(Strain Gauges #2 through #5) place at one point and built-in through the depth of section.

Figure 4.22 points out the locations of each strain gauge through the section. The section

was along the center of span and 14-1/4” to the side of the load.

As shown in Figure 5.54, the top concrete strain at peak load was at a conventional

0.0032, although this value was nowhere near the stress seen in the concrete surface gauges

at failure of the original panels in basically the same location. The strains vary from gauge to

gauge, but the strain variation was not quite linear. The stress in the FRP deck form at peak

load was 0.0114, which, from a linear-elastic relationship translates into a stress of 51 ksi.

This strain was similar to the strains seen in the original three FRP deck forms at failure. For

unknown reason(s), a set of strain gauges (one top and one bottom, #17 and #7 respectively)

set 9” away from the center of load read a much higher concrete strain of 0.09 at peak load,

see Figure 5.55.

0

1

2

3

4

5

6

7

8-0.004 -0.002 0.000 0.002 0.004 0.006 0.008 0.010 0.012 0.014

Strain, in/in

Dep

th o

f Sec

tio

n,in

20 kip 40 kip 60 kip 80 kip100 kip 120 kip 130 kip

Concrete Surface

FRP

Figure 5.54 Strain profile of the five strain gauges, through the depth of section, Panel C2

131

-0.015

-0.010

-0.005

0.000

0.005

0.010

0.015

0.0 20.0 40.0 60.0 80.0 100.0 120.0 140.0

Load, Kips

Str

ain

, in

/in

Top Surface Concrete Bottom Surface FRP

Tension Side

Compression Side

Figure 5.55 Other two Strain gauges 9” from center of load, along center of span, Panel C2

The lateral opening of the shiplap joint was monitored for this test. Figure 5.56

graphs the load versus the lateral displacement for the center shiplap joint. The panel

modifications seem to have increased the initial longitudinal cracking strength. In Panel C,

the lateral displacement started at 3 kips in the inelastic test. This longitudinal concrete

cracking along the shiplap joint probably first developed in the conditioning step, however,

when the panel saw a load of 30 kips. Panel C2 conditioning load of 24 kip was not large

enough to initiate this crack. For Panel C2, this crack did not develop until 47 kip, well into

the inelastic test. Undoubtedly, with concrete strength relatively equal, the added aggregate

and the added FRP grid over the shiplap joint have increased the strength before the

longitudinal cracking starts. A major question remains: which modification was responsible

for the increase in strength? Did the added strength come from the added aggregate or the

FRP Fibergrate Molded Grating reinforcement over the shiplap joint, or a combination

thereof? Panel C2 should have had only one modification change, so a definitive answer to

132

this question could have be deduced. A UW undergraduate student, however, conducted an

independent study to answer this question (Sanders, 2001). According to the results of

testing of this study, the transverse reinforcement capacity over the shiplap joint was

primarily due to the added aggregate to the horizontal surfaces of the FRP deck form and any

capacity as a result of the additional Fibergrate Molded Grating was marginal.

0.0

20.0

40.0

60.0

80.0

100.0

120.0

140.0

0.00 0.05 0.10 0.15 0.20

Displacement, in.

Lo

ad, k

ips

LVDT range exceded at 119 kips

Disp. starts at 47k

Figure 5.56 Lateral displacement of center shiplap joint of Panel C2

Initially, other than the permanent indentation under the load patch, the mode of

failure was hard to definitively define since loading was halted for other reasons. When the

panel was cut the center of the span, the mode of failure was definitively revealed. Figure

5.57 shows the punching shear cracking pattern that was seen in the same post-mortem cut of

Panel C. Again the 45° angle shear cracks ended up at the shiplap joint on the side of the

deck form that, this time around, has aggregate on the outboard side. The outboard side of

the deck form with aggregate cannot transfer vertical shear to the adjacent deck form once

the concrete crack had developed, however, because it lacked a lip overlap or structural

133

continuity. Basically, once the shear crack had developed, the FRP deck form shiplap joints

were only capable of transferring vertical shear to one side, the side with the outboard

possessing the overlap lip. The amount of shear that this lip can transfer was also probably

very limited since: 1.) the lip strength was low, 2.) shear transferred in this manner was

likely to create de-bonding of the adjacent form.

Figure 5.57 Post mortem cut along the center of span for Panel C2 revealed a punching shear

failure

5.6 Accelerated Fatigue Beam Test

The typical goal of an accelerated fatigue beam test is to determine a material’s

resistance capacity under a large number of load repetitions. With bridge decks, design

against fatigue is germane because of the repeated loading delivered from traffic. In addition

to the material’s resistance to repeated loading, this FRP reinforcement system’s strength and

stiffness of its composite section was directly reliant upon maintaining the bond between the

FRP deck forms and the concrete. With these two fatigue concerns in mind, an accelerated

fatigue beam, of size and setup exactly the same as a negative moment beam, was subjected

134

to 2 million cycles of repeated load. Over the course of 2 million cycles the relative stiffness

change was monitored to evaluate effects of fatigue on the beam’s resistance.

As was done for the negative moment beam test, the 200 kip actuator in one span and

the 55 kip actuator in the other loaded the continuous 2-span Beam E2 (fatigue beam). The

200 kip actuator maintained a constant load of 16 kips in one span, while the 55 kip actuator

subjected the other span to a 2-hertz cyclic load that ranged from a low of 4 kips to a peak of

20 kips (20 kips was approximately 28% of the average ultimate loads seen in Beams E). At

every 200,000 cycles the test was stopped to measure deflections at 2 kip load increments up

to 16 kips, while the opposite span was held at a constant 16 kip load. The cyclic loads were

selected, somewhat arbitrarily, on the basis that their range was equal to the prescribed loads

of the AASHTO designated wheel for the HS20-44 truck.

Figure 5.58 Preload concrete cracks for the fatigue beam

Some cracking occurred as the test specimen was moved to the lab and positioned for

testing. All concrete cracks that existed before the test started were marked and monitored,

135

see Figure 5.58. No instruments were applied to the fatigue beam and therefore the

deflections were determined with stroke displacements. After the accelerated fatigue test

was complete, the beam was then subjected to an inelastic test, in which the procedure used

for the negative moment beam test was followed.

02000400060008000

1000012000

1400016000

18000

0 0.02 0.04 0.06 0.08 0.1 0.12Stroke Deflection, in

Load

, pou

nds

initial 200000 600000 1000000 1400000 1800000 2000000

Kinitial = 255 kip/in

Figure 5.59 Relative stiffness of the 200-kip span during the fatigue test after

various number of cycles

Figure 5.59 graphs the relative stroke displacement measured at 400,000 cycle

intervals, starting after 200,000 cycles. The measurement was the deflection in the 200-kip

actuator span under a 16 kip load while the 55-kip actuator span held a constant 16 kip load.

Counter intuitively and for an unknown reason, the initial load measurement showed a lower

stiffness than the interval measurements. Consistently for each interval of measurements,

there was a relatively low stiffness for the first 2 kips. Then, as load increased so did the

stiffness increased and over 2 million cycles, a initial stiffness of about 255 kips per inch was

maintained (initial stiffness is defined as a secant line to along the established load vs.

deflection slope up to 16 kips).

136

Looking back at Beam E1 and E3 tests, the tensile strain in the FRP bi-directional

grid was low because they were uncracked sections over the middle support (with 16 kip

loads in each span), thus the top concrete surface was carrying the tension. The fatigue beam

was cracked over the support before the test ever started. For the fatigue beam, a cracked

section over the middle support was essential to ensure the FRP grid was providing the

tensile capacity over the support and, therefore exposed to tensile fatigue stress. There was

very little change in the fatigue beam stiffness over 2 million cycles. There was no growth in

existing crack lengths or any new crack development over 2 million cycles.

02000400060008000

1000012000140001600018000

0 0.02 0.04 0.06 0.08 0.1 0.12

Displacement, in.

Lo

ad, p

ou

nd

s

Initial 200000 600000 1000000 1400000 1800000 2000000

Kinitial = 250 k/in

Figure 5.60 55 kip actuator relative stiffness of 2 million cycle fatigue test

The relative stiffness change for the span with the 55 kip actuator over 2 million

cycles is exhibited in Figure 5.60. There was no noticeable change in the slopes of the

200,000 cycle lines, or stiffness, over 2 million cycles. The discrepancy between the initial

measurement and the other measurements within the first 200,000 cycles can be attributed to

two things: 1.) a small loss in stiffness up to a load of 8 kips, 2.) the disintegration of the

137

plaster layer between the 3” square bar and the top surface of the beam. Once this layer was

gone, there was a natural shift in the relative load versus displacement measurements. In

hindsight, a more durable connection between the actuator and the beam in the span exposed

to the cycling load could have prevented this shift.

Another step in evaluating the effect of fatigue on the beam was to perform an

inelastic test after the beam had been subjected to the 2 million cycles. Since it had the same

physical characteristics and was from the same pour as the negative moment beams, a

comparison of the ultimate load of a fatigue beam (exposed to 2 million cycles of a wheel

load) with the ultimate load of the “virgin” negative moment beams should provide an

indication of relative loss of ultimate strength and stiffness. The inelastic test was conducted

according to the protocol carried out for the previous negative moment beam tests.

0

10

20

30

40

50

60

70

80

0 0.5 1 1.5 2 2.5Center span LVDT deflection, in

Lo

ad, k

ips

Span A Span B

P yield = 59 k

K eff = 236 k/in

Figure 5.61 Load versus deflection curves for Span A and Span B for Fatigue Beam

During the inelastic test, unknowingly data was not recorded for the 55 kip actuator to

a load of 25 kip when it was noticed and fixed. Therefore, there was data missing for the

138

first 25 kips for the 55 kip actuator span, Span B. At 37 kips, the first new crack developed

at the edge of the bearing pad on the middle support on the side of Span B. At 58 kips a new

cracked formed at the edge of the bearing pad on the middle support on the side of Span A.

At 65 kips a large single flexure crack developed in Span A on the side of the beam without

bond between the FRP deck form and the concrete, see Figure 5.62B. At 67 kips and at

0.42” deflection of Span A, two large shear cracks failed the beam on one side of Span A, see

Figure 5.62A. The test continued until the concrete crushed under the 3” square bar under the

200 kip actuator in Span A.

As seen in Figure 5.61, the initial stiffness for the initial cycle of Span A of the

fatigue beam was 250 kips per inch, where, as illustrated in Figure 5.61, after over 2 million

cycles the effective stiffness for the fatigue beam was 236 kips per inch. The ultimate

strength was 68 kip and the yield point was 59 kips. The effective stiffness and the yield

were slightly lower than Beams E1 and E3, which did not see 2 million cycles before they

were tested to failure. From a strength perspective, however, the failure load of Beam E2

(Fatigue) was very consistent with Beams E1 and E3.

139

Shear Cracks

A B

Figure 5.62 Cracking on each side from the inelastic test of fatigue beam

Failure of the fatigue beam during the inelastic test was due to the two shear cracks

(on the side of the beam with the bond between the FRP deck form and the concrete) in

Figure 5.62A, just like shear cracks seen in the failures of the negative moment beam tests.

5.7 Concrete Cylinder Strengths

Three concrete cylinders were made for each pour and were tested according to

ASTM C-39 to determine the 28-day unconfined compressive concrete strength. The

concrete cylinders were covered and stored in the same environment that the specimens were

kept. On the day of the inelastic test of each specimen, an additional three cylinders were

tested to record the concrete compressive strength of the day of the test. The results are

tabulated in Table 5.3. The 4 ksi, Wisconsin Department of Transportation Class D, concrete

mix ingredients are tabulated in Appendix B.

140

Table 5.3 Concrete strengths (± one standard deviation) tested according to ASTM C-39 Specimen 28-Day Concrete

Strength, psi (WisDOT Class D)

Concrete Strength on Day

of Test, psi

Days Between Pour and Test

Panel A 5537 ± 290 5367 ± 20 45 Panel B 5537 ± 290 6200 ± 95 59 Panel C 5537 ± 290 5860 ± 240 73 Panel C2 5303 ± 320 5303 ± 320 28 Beam D1 6308 ± 190 6343 ± 240 87 Beam D2 6308 ± 190 7097 ± 150 86 Beam D3 6308 ± 190 7140 ± 90 95 Beam E1 6308 ± 190 5857 ± 645 116 Beam E3 6308 ± 190 4248 ± 250 121 Beam E5 6308 ± 190 6473 ± 195 109 Fatigue Beam 6308 ± 190 5880 ± 90 192

141

6 DATA ANALYSIS OF TEST RESULTS

6.1 Introduction

The next step was to analyze and refine the experimental data into informative results

and ultimately fulfill the stated objectives. As the previous chapters, the data analysis will be

presented and laid out according to the sequential order of group testing. One exception is

the inclusion of the test results from the additional concrete deck panel, Panel C2, with the

original three concrete deck panels. Taken as a whole, the main purpose of this chapter was

to evaluate the behavior, based on the objectives, of the FRP reinforcement system’s

structural performance. Part of fulfilling the main purpose was to take the results from this

unconventional reinforcement system and “stack them up” against selected strength and

serviceability requirements established for conventional steel reinforced concrete decks. An

additional goal that will form the purpose of the next chapter was to develop some design

guidelines from the experimental data and analytic simulations of the FRP reinforced deck

system.

In the course of analyzing the data results presented in Chapter 5, assumptions will

periodically be made and described. This was a key task when converting the loads applied

to the panels and beams into moments. Some of the experimental results will be compared to

pre-testing analytical predictions (derived from existing design equations or AASHTO design

methods) that are also based upon some given assumptions.

142

6.2 Concrete Deck Panels

As has been stated before, there were three main objectives for the concrete panel

tests. First, is to attempt to determine the load distribution widths at service and ultimate

loads for a defined 9’ wide concrete deck panel. The experimental load distribution widths

will then be compared to an AASHTO equation used to determine load distribution in

conventional concrete decks. The effect of the shiplap joints and the lack of distribution steel

might be found by comparing the results from the experimental distribution widths to the

AASHTO equation. Second, was to determine the mode of failure. The visual and data

results clearly identify the mode of failure for all of the concrete deck panels as punching

shear. These results will be compared to an ACI 318 equation for punching shear. Finally,

the ultimate strengths need to be compared to the service load to derive factors of safety.

Since failures are based on punching shear capacity, the ultimate capacity will generally be

independent of boundary conditions or span length (for normal deck span ranges).

6.2.1 Positive Moment Distribution Width

As well as serving as the top flange of the longitudinal girders, another major

function of a concrete deck is to distribute the live load to the girders. Live loads applied on

the concrete deck are distributed in a lateral direction as the load “travels” to the supporting

girders. The distribution width of the concrete deck that supports the live load centered in

the span between the girders is usually assumed to be a function of the clear span between

the bearing edges on the top flange of the parallel girders. Maximum positive moment load

distribution is expected to occur at the girder support. Another words, as the load or moment

travels toward the girder, from the load, it also travels laterally, picking up added section for

143

capacity. Following this logic, the minimum load distribution is expected at the point of live

load application. The minimum design distribution width at the center of span is the

distribution width required to resist the positive design moment produced by a wheel load,

assuming that the moment is constant in the portion of the deck under the distribution width

and equal to the actual peak deck internal moment.

To determine the positive moment distribution width from the concrete deck panel

test results, experimental strains on top and bottom at various points along the centerline of

the span are used to determine the moments based on the moment-curvature relationship

measured in the positive moment beam experiments.

0

500

1000

1500

2000

2500

0 0.0005 0.001 0.0015 0.002 0.0025 0.003

Curvature(radians)

Mo

men

t, in

-k

D1 D2 D3, No top grid

0.00091 0.0016

See Table 6.1

0.00188

1650

Figure 6.1 Moment-curvature relationship for the positive moment beams

The positive moment beam test results are presented in the form of moment-curvature

relationships and plotted in Figure 6.1. They are needed to define the moment distribution

curve of the concrete deck panels. It is assumed that the selfweight strains, which are not

measured, can be ignored or are similar in the “D” beam and panel tests. Therefore, only the

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applied load is producing the strains and thus, the moments. In order to determine the

moments in the positive moment beam test, static analysis is used to calculate the mid-span

moment for a simply supported beam. The static moment Equation 6.1, is applicable.

M = PL/4, L = 104” (6.1)

An average of the three beams was used to derive a linear relationship for moment-

curvature up to a moment of 1000 kip-inches. At this moment, the moment-curvature

relationship is assumed to change based on the visible slope in Figure 6.1. A second slope

and another linear relationship is derived between 1000 and 1500 kip-inches. Table 6.1

shows the measured curvatures for each beam at the assumed positive moments. The linear

relationship up to 1000 kip-inches will be applied for curvatures less than that. The linear

relationship above 1000 kip-inch moment will be used up to 1500 kip-inches.

Table 6.1 Table of normalized curvature values to establish an approximate linear relationship

Curvature, θ (radians) Beams D Load, kip

Moment, kip-in For a 36” width

D1 D2 D3 Average Moment/

(Ave*36”) (Kip-in/radian

per inch width) 38.5 1000 0.000843 0.000979 0.000917 0.00091 30433 57.7 1500 0.001629 0.001675 0.001503 0.00160 20142 63.5 1650 0.002019 0.0018623 0.001760 0.00188 14981

When the concrete deck panel moments for a 20.8 kip wheel load (HS20 wheel load

with impact) are graphed along the center span, see Figure 6.2, a bell shaped curve typically

represents the moment distribution per inch of width from the center of span for a 9’ wide

concrete deck panel. The profile of the curves through the center of the load was

interpolated, since data in this area was not available. Also, the curves were extrapolated

beyond the last point of measurement nearest the panel side to the panel side. These two

145

assumptions were necessary to integrate the area under the curves for the calculation of the

distribution width for a 9’ wide concrete deck panel.

Theoretically numerically integrating the area under the moment distribution curve

and dividing this summation by the estimated maximum moment experienced at the location

of the live load can determine the distribution width. The area under the curve should be

integrated from where the moment is zero on one side of the load to a zero moment on the

other side of the load. However, the experimental concrete deck panels were a finite width

and the values from the experimental results represent a distribution width for a 9’ wide piece

of deck. Therefore, the area under the curve cannot be greater than and must be equal to the

static moment, Equation 6.1, applied so the theoretical integration of the area under the curve

is not necessary. Instead, for the finite width of the panel Equation 6.2 can calculate the

distribution width. If the width were longer, as in the prototype bridge (as in the width in the

direction of traffic), the moment in the tails of the curve would be lower but the tails would

extend out further. The peak moment under the load would be nearly the same.

0123456

-54 -36 -18 0 18 36 54Distance from center of load, in

Mo

men

t, k

ip-i

n p

er i

nch

wid

th

Panel C Panel C2

Mmax

Figure 6.2 Positive moment distribution for the elastic load of 20.8 kips for a 9’ wide panel

Distribution Width = Area of Static Moment/ Mmax = (PL/4)/Mmax (6.2)

146

Since data was not recorded for the first 36 kips for Panel A and the loss of accuracy

of the strain gauge data for Panel B, the data in the elastic region for these panels were not

available or not of good quality. Hence, the moment distributions for these panels are not

plotted or calculated for the elastic load, 20.8 kips.

0

10

20

30

40

50

60

-54 -36 -18 0 18 36 54Distance from center of load, inches

Mo

men

t, k

ip-i

n. p

er in

ch w

idth

Panel A, 96 kip Panel B, 87 kipPanel C, 91 kip Panel C2, 130 kip

Figure 6.3 Positive moment distribution at panel failure

Figure 6.3 depicts the moment distribution per inch of width of the concrete deck

panels at their ultimate loads, respectively. Table 6.2 gives the calculated distribution widths

for the moment-distribution curves shown above relative to AASHTO’s equation 3.24.3.2,

Case B, Method 1, used to determine distribution width (see equation 6.3). In essence, this

method takes a truck wheel load and divides it by the distribution width to derive a live load

moment per foot of width for a slab supported on an abutment or piers. Comparing the

experimental values shows that the values for all panels were greater than that of AASHTO.

The ultimate distribution widths range between 5 and 18% greater than AASHTO’s equation.

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The service load distribution widths for Panel C and C2 are substantially greater than

AASHTO.

Table 6.2 Live load distribution widths for the 9’ wide concrete deck panels

Live Load Distribution Width, inches Load, kip Panel A, 11’-6” Panel B, 9’-10” Panel C, 8’-0” Panel C2, 8’ 0” 20.8, HS20 w/impact n/a n/a 94 101

Ultimate 66 58 62 62 AASHTO, eqn. 6.3 56 55 54 54 Ultimate/AASHTO 1.18 1.05 1.15 1.15

E = 4 + 0.06*S, S = span (ft) (6.3)

On this basis of the test results and the data analysis, as well as knowing that the

distribution width for an infinitely wide concrete deck panel would be equal to or greater

than the experimental results for a 9’ wide panel, the equation for AASHTO distribution

width used for steel reinforced decks, equation 6.3, is applicable as a design guide for the

FRP reinforcement system.

The distribution test results may now answer the issue of longitudinal (transverse to

the direction of traffic) concrete cracking along the shiplap joint effect upon the ability to

distribute the load laterally (or longitudinally in the context of a actual bridge deck/girder

system). The presence of the shiplap joint, and the lack of distribution reinforcement on the

bottom portion of the concrete deck, creates a tendency for developing these longitudinal

cracks. A longitudinal concrete crack along the shiplap joint is a structural discontinuity that

reduces the ability to distribute load laterally by preventing moment transfer laterally. For

Panels B, C and C2, the center shiplap joint was monitored for lateral displacement, which is

a measure of the degree of longitudinal concrete cracking above the shiplap joint.

148

Longitudinal concrete cracking along the center shiplap joint (directly centered under the

load area) would have some impact on the distribution width. What was a concern in the

laboratory testing was the possibility of longitudinal concrete cracking at the first shiplap

joint clear of the load area. In a bridge application, the dynamic live loads will travel over all

of the shiplap joints. If the actual bridge deck is loaded sufficiently to cause longitudinal

concrete cracking at each shiplap joint, the distribution width will be diminished.

In Chapter 5, the lateral opening of the center shiplap joints were graphed for Panels

B, C and C2 in Figures 5.14, 5.22, and 5.56, respectively. In each test, the center shiplap

joints opened significantly by the time ultimate was reached. Panel C2, (which had the

added aggregate to the FRP deck form) however, showed a dramatic increase in resistance to

the load that initiated crack growth, from 5 kip to 47 kip (Panel C and C2, respectively).

This improvement must be linked to the modifications to the reinforcement scheme.

Assuming the Fibergrate Molded Grating does not contribute to the increase in transverse

reinforcement capacity across the shiplap joint (as concluded from Sanders, 2001), the added

aggregate could prevent the development of concrete cracking along the shiplap joint in an

actual bridge, however, this should be investigated further.

6.2.2 Punching Shear Capacity

Although knowing the distribution width is important to the protocol of concrete

bridge deck design, from a practical point of view it may not affect the ultimate capacity of

most concrete bridge decks. As a consequence of a stiff continuous concrete plate, punching

shear usually occurs before a bending failure is achieved. For a normal deck design of a

slab-girder system, however, AASHTO does not require design check for punching shear.

149

Since AASHTO does not require a design check, there is not a punching shear equation

specified by AASHTO. Therefore, the punching shear capacity of the concrete deck panels

should be compared to American Concrete Institute punching shear equation in ACI 318

building code requirements.

For a given deck thickness and concrete strength and since punching shear capacity is

independent of the span length, one value could apply to all panel. Therefore, with the load

area and the depth of section equal for each panel, the punching shear capacity was a

function of the concrete strength on the day of the respective test.

Vc = 4 * (f’c)1/2 * bo * d (6.4)

Equation 6.4 is the ACI equation used to calculate the nominal punching shear

capacity for rectangular loaded areas, which was the case for the tested concrete deck panels.

The variable bo is the critical perimeter section of the rectangular loaded area. ACI defines

bo as “…section located a distance d/2 from the periphery of the concentrated load.” It was

assumed that the load at ultimate is covering the area of the steel plate, 8” x 23”, on top of the

elastomeric pad. Figure 6.4 depicts the loaded region applied by the 8” x 23” steel plate on

top of the elastomeric pad on the surface of the concrete deck panel. Table 6.3 compares the

experimental values of ultimate loads to the calculated punching shear capacities based on

equation 6.4.

150

14.93" 9' Width

29.93"

d = 6.93"

Span Varies

Sim

ple

Supp

ort

Sim

ple

Supp

ort

Critical Load Perimeter

Actual Load Perimeter

Figure 6.4 Load perimeter area to define bo

Panel f’c on the

Day of Testing

psi

Experimental Punching

Shear, kips

Punching Shear

Capacity based on

ACI 318, kip d = 6.93”

Punching Shear

Capacity based on

ACI 318, kip d = 5”

Experimental Shear / ACI 318

Capacity based on d = 6.93”

Experimental Shear / ACI 318

Capacity based on d = 5”

Panel A 5370 96 131 107 0.73 0.90 Panel B 6200 87 141 115 0.62 0.76 Panel C 5860 91 137 111 0.66 0.82 Panel C2 5303 130 130 106 1.00 1.23

Table 6.3 Punching shear values and calculated capacities for the concrete deck panels

The ACI equation, Equation 6.4, is given for steel reinforced concrete. This equation

may or may not be applicable to the current deck because of the unconventional issues of the

shiplap joint, partial areas of deck forms with bonded aggregate, and the use of stay-in-place

FRP deck forms that act as the reinforcing material. It is quite clear from the values in Table

6.3 that the unconventional issues of the FRP reinforcement system may reduce the punching

shear capacity of the panels with the FRP deck forms without the added aggregate coverage.

As seen by the ultimate capacity increase between Panel C and C2, the increased coverage of

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aggregate increased punching shear capacity by 43% (assuming the added Fibergrate

Molded Grating modification above the shiplap joints of Panel C2 was inconsequential to the

increased punching shear capacity). Panel C2 seems to meet and slightly surpass the ACI

equation of punching shear capacity. Therefore, with the added aggregate coverage on the

FRP deck form, the ACI equation may be considered applicable to calculate punching shear

capacity.

Table 6.3 has two values for the punching shear capacities depending upon the

effective depth of the cross section. Evaluating the significant gaps between the

experimental and the ACI punching shear capacity values, as well as, looking back at Figure

5.23 post mortem of the Panel C, it would be incorrect to assume an effective depth of the

tensile reinforcement at the centroid of the cross section of the FRP deck form (with respect

to calculating punching shear capacity). As can be seen in Figure 5.23, when the shear crack

reached the top of a corrugation on the left hand side of the figure, it traveled along the

smooth surface of the FRP deck form to the “lipless” edge of a shiplap joint. This portion of

the FRP had no aggregate attached and little bond capacity. This effectively reduced the

effective depth of section, d, used in the punching shear capacity equation to represent the

total depth of section to the top of the corrugation or 5”. To stay consistent with the

definition of the ACI equation for the effective depth, however, the centroidal representation

for “d” would be used. The 5” depth brings predicted capacities closer to the measured

values and may be more appropriate, except for Panel C2 where additional surfaces of the

FRP deck form had aggregate coating and higher bond capacity.

To use the ACI equation for determining the punching shear capacity for the FRP

reinforcement system without the added aggregate on the FRP deck forms, an empirical

152

coefficient must be used to reduce the capacity of equation 6.4 to represent the experimental

values. The coefficient should be equal to or below that of the lowest ratio of the

experimental strength to ACI punching shear capacity, see Table 6.3. Thus for the

incomplete coverage of aggregate to all horizontal surfaces of the FRP deck form and

assuming the ACI definition for “d” (6.93”), the coefficient should be 0.60 to represent the

lowest strength to capacity ratio for Panel B. For the complete coverage of aggregate to all

horizontal surfaces, test results from Panel C2 indicate that no coefficient is required.

6.2.3 Evaluation of Ultimate Capacity and Effective Stiffness

From a safety point of view, the most important design requirement is ultimate

strength. As mentioned previously, each deck panel failed by punching shear. From a

serviceability point of view, the effective stiffness of the section is important to satisfy

requirements for deflection. The effective stiffness result was analyzed and used for relative

comparisons, but other than the quick calculations done in Chapter 5, the deflections from the

experimental tests are not apart of the scope and will not be formally addressed. However, at

the end of this Chapter, there is a table of deflection and stiffness values of each specimen to

compare performance.

Table 6.4 Strength and effective stiffness results of the concrete deck panel tests

Panel Span, ft

Effective Stiffness*,

Kips/in

Yield Load*, Kips

Ultimate Capacity,

Kips

Factor of Safety, Ultimate load/

Service load of 16kip A 11’-6” 88.1 80 96 6.0 B 9’-10” 152 77 87 5.4 C 8’-0” 153 80 91 5.7 C2 8’-0” 217 106 130 8.1

* Priestley method, explained in Appendix D

153

Although not linear, logically the stiffness shown in Table 6.4 increases as the span

length decreases. In addition, the longitudinal strength and stiffness both increase in Panel

C2 due to the modifications made. Assuming the Fibergrate Molded Grating placed over

the shiplap joints is ineffective or insignificant in increasing the longitudinal bending

stiffness, the strength and stiffness in Panel C2, over Panel C, increased by 43% and 42%,

respectively due to the increased coverage of the aggregate on the FRP deck form. If the

panels are normalized, by ratio of the concrete strength values, then the strength and stiffness

increase by 49% each. In addition, the factors of safety for strength are sufficient, especially

with the reinforcement modifications.

6.3 Positive Moment Beams

The results of the positive moment beam test will not have any direct impacts on

establishing design recommendations, such as distribution width or ultimate capacity of an

actual bridge deck. Instead, the results from the positive moment beam testing are used

indirectly. For example, one main use of this data is as a tool to associate positive moments

to curvature results from pure bending. In turn, this moment curvature relationship is used to

translate the curvatures measured from the concrete deck panels into bending moments, like

what was done earlier for determining the distribution width. In addition, another main

purpose of the positive moment beam is to assess the structural behavior of the FRP

reinforcement system in a pure bending environment, such as strength per inch of width for

an 8’-8” span.

Also aiding hard quantitative results of the positive moment test are qualitative

observations. An example of these observations was the subtle lesson of partial composite

154

behavior due to the limited horizontal shear transfer between the FRP deck form and the

concrete. This limited shear transfer was caused by the incomplete coverage of the aggregate

on the horizontal surfaces of the FRP deck form. The partially composite behavior would not

have been easily detected from the concrete deck panel test alone. Another goal of these

tests was to examine the positive moment behavior without the top bi-directional FRP grid

reinforcement. Without the top grid, the Beam D showed a lower post-ultimate capacity.

6.3.1 Positive Moment Curvature Relationship

Some initial data analysis from the positive moment beams has already been

presented earlier as part of the methodology to determine the concrete panel distribution

widths, specifically in Section 6.2.1. In general, the group of three positive beam moment

curvature curves does exhibit very similar behavioral trends, thus providing, at least

consistent repeatability of testing results. This consistent repeatability of the testing results

does help breed confidence in the integrity of the test data.

6.3.2 Evaluation of Load versus Deflection, Ultimate Capacity and Effective Stiffness

The load versus deflection relationship in the primary direction for all three positive

moment beam tests is plotted as a group in Figure 6.5. There was a general agreement

between the shapes of the curves, especially in the elastic region. All beams had very similar

behavior up to a load of 40 kips and they all experienced a subtle change in initial stiffness at

a load in the early to mid 30 kip range. After that, Beams D2 and D3 followed the same path

up to and just through their respective ultimate loads. All of the beams reached ultimate

capacity at a similar deflection of one inch. Assuming equation 6.1 is applicable, the positive

moment bending strength per inch of width for an 8’-8” span is calculated for each of the

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three-foot wide beams in Table 6.5. Tabulating the results for all three positive moment

beams the average strength per linear inch is 52.33 kip-inches, which translated into a load of

2 kip for a one inch wide strip on an 8’-8” span.

0

10

20

30

40

50

60

70

80

90

0 0.5 1 1.5 2 2.5 3 3.5 4Deflection, inches

Lo

ad, K

ip

D1 D2 D3 (no top grid) Figure 6.5 Load versus deflection curves for all of the positive moment beams

Positive Moment Beam

8’-8” Span

Effective Stiffness* (Kips/in)

Yield Load* (Kips)

Ultimate Capacity (Kips)

Moment Capacity

Per Inch of Width

(Kip-in)

Stress in FRP Deck Form at Beam Failure (Ksi)

Mode of Failure

D1 118 54 63.5 45.8 43.5 Flexure/Concrete Crushing

D2 102 66 78 56.3 54.5 Same as above D3 (no top grid)

126 68 76 54.9 48.4 Same as above

* Priestley method, explained in Appendix D

Table 6.5 Strength and effective stiffness results of the positive moment beams

As mentioned in Chapter 5 for the test results for Beam D3, comparing the post-

ultimate behavior of Beams D1 and D2 to D3 provided a conclusion that the top FRP bi-

directional grid did exhibit a role in providing post-ultimate bending capacity once the

156

concrete crushed. This post-ultimate compression capacity allowed the beam to carry a

significant load through large deflections, which is valuable from a safety point of view.

From an elastic behavior point of view, comparing the results between the beams with and

the beam without the top grid, the beam without the top grid had a greater initial effective

stiffness. Without additional testing, however, it is difficult to make any general conclusions

about this difference in stiffness and the role the top grid may have had to play.

6.4 Negative Moment Beams

Like the positive moment beam tests, the results from the negative moment beam test

do not directly derive design recommendations. Indirectly, the results from the negative

moment beam test, combined with analytical modeling, were used as a guide to determine a

negative moment distribution width. The specific data of interest are the loads in the early

elastic range that helped determine the concrete cracking moment in the negative moment

region over the girder. The importance of determining the negative moment distribution

width is more of a serviceability issue rather than strength. It has been determined that

punching shear was the mode of failure, so from the results of the laboratory testing, the

negative moment strength is not a significant design issue. In addition, the negative moment

beams failed in shear; hence, the negative moment capacity could not be determined. The

serviceability concern is the load that will cause concrete cracking over the girder. Some

general negative moment beam behavior, not directly involved with the negative moment

distribution width, was documented, such as the negative moment curvature relationship.

157

6.4.1 Negative Moment Curvature Relationship

To simplify the analysis, an assumption was made that the two span negative moment

beam was pin supported at one end with the middle and opposite ends as roller supports. The

assumption was reasonable for the narrow plaster bearing pads that supported the beam. It

was also assumed that the concentrated loads were perfectly centered in their spans and the

two span lengths were identical. These assumptions are accurate if dimensional tolerance of

±0.5” is accepted. The preceding assumptions facilitated a simplified negative moment

calculation over the middle support. Equation 6.5 is the moment equation, neglecting

selfweight, used to determine the negative moment over the support. The selfweight moment

was small and neglected. The data acquisition device did not measure the selfweight strains.

Therefore, the only load that was producing the strains and thus, the moments was from the

actuators. Equation 6.5 is only valid for loads (P) up to 50 kip, when the loads in each span

were still equal, which worked well since this analysis was only concerned with loads in the

early to mid elastic region.

M = 0.187*P*L, L = 96” (6.5)

Figure 6.6 shows the relationship, up to 50 kip load in each span, of equation 6.5 to

calculate the moment centered over the middle support versus the curvature (defined by the

interior FRP top grid strain gauge and the bottom surface concrete strain gauge) of the beam

at the same location.

158

0.0100.0200.0300.0400.0500.0600.0700.0800.0900.0

1000.0

0 0.0005 0.001 0.0015 0.002 0.0025 0.003Curvature, radians

Mo

men

t, in

-kip

s

E1 E3

Figure 6.6 Negative moment curvature relationship centered over the middle support, Beams E

The moment curvature relationship was established with data from only two beams,

since the third beam was not equipped with an internal strain gauge on the FRP top grid.

Both beams exhibited a significant change in slope of the moment curvature curves between

the moments of 200 to 400 inch kips. This range in the change of slopes is thought to

represent the top surface of the concrete cracking over the middle support and transferring

the tension resistance to the top FRP bi-directional grid alone. The cracking moment

equation defined by ACI 318 (and AASHTO) is shown in equation 6.6.

Mcr = fr*Ig / yt (6.6)

Where the modulus of rupture,

fr = 7.5*(f’c)1/2 (6.7)

With a transformed gross section of 720 in^4 for and 18” width (see Appendix D for

section properties), yt = 3.96” and an average f’c = 5052 psi, the cracking moment, Mcr, is 97

inch kips per 18” of width or 194 inch kip per negative moment beam. This is equivalent to a

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10.8 kip load in each span. As expected, the analytical cracking moment calculated above is

close to the moments observed when the slopes of the moment curvature curves started to

change. Working backwards with equation 6.5, this time including selfweight (0.125*w*L2),

the applied load to cause this cracking would be 10.3 kips in each span.

The experimental loads that caused the initial concrete crack over the middle support

for Beams E1 and E3 were 7.8 and 12 kip, respectively. The average experimental cracking

load for of Beams E1 and E3 is 9.9 kip, which is slightly lower than the calculated load of

10.3 kips. The loads to cause cracking in the negative moment beams were for an internal

relative comparison between beams. The general agreement between the experimental and

analytical cracking moment values, however, was a crucial key for verifying the calculated

cross-sectional properties used to analytically model the negative moment distribution width

of the prototype bridge system in Chapter 7.

6.4.2 Evaluation of Load versus Deflection, Ultimate Capacity and Effective Stiffness

The load versus the center deflection relationships of Span A LVDT for the negative

moment beam tests (except Beam E5) and the fatigue beam are graphed in Figure 6.7. There

was general agreement between the load versus deflection relationships for the beams,

including the fatigue beam (after seeing 2 million cycles of load). The general agreement

between the negative moment and fatigue beams also suggests that the FRP system did not

see any significant loss in strength after being exposed to 2 million cycles.

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01020304050607080

0 0.5 1 1.5 2Center Deflection, 200 kip Span, inches

Load

, Kip

E1 E3 E2, Fatigue

Figure 6.7 Load versus deflection curves for Span A LVDT for the negative moment beams

Table 6.6 documents further evidence to the conclusion stated in the preceding

paragraph. Table 6.6 shows the initial stiffness, mode of failure, yield and ultimate loads of

Span A of all the negative moment and the fatigue beams. The effective stiffness, yield and

ultimate load values for the beams were very close to each other. The initial stiffness of

Beam E1 and the E3, 249 and 272 kips per inch, respectively, were slightly higher than Beam

E2 (Fatigue Beam). The slight drop off in the effective stiffness and the yield load in Beam

E2 may indicate a slight degradation in stiffness and yield over the fatigue loading. This,

however, was not a significant drop and thus does not draw concern about practical

performance of the FRP reinforcement system in an actual bridge because a loss of some

stiffness in steel reinforced bridge decks exposed to 2 million cycles would be expected.

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Table 6.6 Strength and effective stiffness results of Span A for the negative moment beams

Negative Moment Beam

8’-0” Spans

Effective Stiffness* (Kips/in)

Yield Load* (Kips)

Ultimate Capacity (Kips)

Stress in FRP Top Grid at

Beam Failure (Ksi)

Mode of Failure

E1 249 66 73 60 Shear E3 272 61 67.5 53 Shear E5 n/a n/a 72 n/a Shear E2 (Fatigue) 236 59 68 n/a Shear

* Priestly method, explained in Appendix D

6.5 Accelerated Fatigue Beam

In general, the test results of the accelerated beam, Beam E2, test lead to the

conclusion that there is no concern for fatigue failure of either the material or the mechanical

bond between the concrete and the FRP deck forms. The testing results documented in

Section 5.6 and Figures 5.59 and 5.60 reveal no significant stiffness loss, for service wheel

load, over 2 million cycles. The maximum load of the cycle was arbitrarily set, 20.8 kips and

the 16 kip range was chosen to represent AASHTO HS-20 truck without impact. The

maximum and amplitude loads were large enough to be considered to induce a significant

non-reversal stress for elastic load levels. Assuming a linear elastic relationship in the early

elastic range with a cracked moment of inertia of 135.2 in^4, yt = 4.50” and negative moment

of 360 kip inches (including selfweight) at the worst loading of each cycle (20.8 kips in one

span and 16 kip in the other) the stress in the FRP grid can be determined by,

fgrid = M20.8*yt / Icr, (6.8)

The maximum stress experienced in the grid per cycle is about 12 ksi, which is about 14.4%

of the tensile strength for the bi-directional grid. One conclusion is that the fatigue life of the

FRP grid must be higher than 12 ksi for 2 million cycles. Looking back, this stress level

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could have been increased. However, for a bridge application with loading in the early

elastic range the grid should not experience any problems with fatigue strength.

The other fatigue issue dealt with the bond fatigue strength of the mechanical shear

transfer between the FRP deck form and the concrete, which is germane for composite

action. In the absence of possible impacts from environmental issues, there is no evidence

from the fatigue testing to suggest that there will be any fatigue strength issues with this

bond.

In addition to performing adequately for fatigue, Table 6.6 shows the fatigue beam

provided equal strength capacities from the inelastic test, as the other 3 negative moment

beams, even after the fatigue beam was subjected to 2 million cycles.

6.6 Performance of Test Specimens Based on Stiffness

This section goes beyond the scope of this thesis, however, it was conducted to

provide some comparison between the measured and calculated stiffnesses of the specimens.

Table 6.7 summarizes the performance of the FRP reinforced test specimens by comparing

the span-to-deflection ratios (i.e., L/800) at a service wheel load, of 16 kips, experienced

during each respective inelastic test (selfweight not included). Typically in bridge design,

the maximum allowable deflection includes impact loading, however, these specimens do not

directly represent an actual bridge member. For example, the distribution widths of the

panels and the end conditions of all specimens were not representative of an actual

slab/girder bridge deck. Therefore, comparing the values in Table 6.7 to AASHTO criteria

was not applicable. Instead, these values are for relative comparison of performance and the

HS44-20 service wheel load (without impact) was selected somewhat arbitrarily.

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Table 6.7 Span-to-deflection ratios for the tested FRP specimens

Specimen Span, Center-to-Center of Bearing

Center Span Deflection at 16

kips, in

L/ (Span/Deflection)

Panel A 11’-6” 0.087* L / 1586 Panel B 9’-10” 0.019 L / 6210 Panel C 8’-0” 0.027 L / 3556 Panel C2 8’-0” 0.054 L / 1778 Beam D1 8’-8” 0.068 L / 1529 Beam D2 8’-8” 0.112 L / 929 Beam D3 8’-8” 0.096 L / 1083 Beam E1 8’-0” 0.038 L / 2526 Beam E3 8’-0” 0.042 L / 2286 Beam E5 8’-0” not available --- Beam E2 (Fatigue Beam)

8’-0” 0.039 L / 2462

* Data was interpolated because data was not recorded until 36 kips into inelastic test Table 6.8 compares calculated stiffness based on gross transformed and cracked

moments of inertia to the initial (based on secant line to 16 kips) and effective (based on the

Priestley Method) stiffnesses measured during the inelastic tests, respectively. The

calculated stiffnesses for the panel specimens were based on the AASHTO positive moment

distribution widths, equation 6.3. The calculated stiffness for the beams was based on the

full width, 3’, for each specimen. The measured initial stiffness and calculated stiffness,

based on gross transformed, were compared at a 16 kip load. The measured effective

stiffness and calculated cracked stiffness were compared at the same load used to determine

the measured effective stiffness using the Priestley Method, 75% of the peak load. The

values for the longitudinal gross transformed and the cracked moments of inertia based on

ACI 440H are located in Appendix D.

Basic deflection equation, PL3/48EI, was used to determine the deflections for the

positive moment simply supported panels and beams with concentrated load at center of

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span, P, and the moment of inertia, I (gross transformed or cracked). To find the deflection

of the 2-span continuous negative moment beams, a computer analysis was conducted for 50

kip concentrated load in each span, which was close to the 75% of the ultimate load seen in

each of Beams E. Again the gross transformed and cracked moments of inertia were

accounted for in the computer analysis. The modulus of elasticity, Ec, for each specimen was

determined from the compression strengths found from the concrete cylinder tests conduced

on the day of the test and using ACI 318 equation, 57,000*(f’c)1/2. It should be noted that

the calculated cracked stiffness, Kcracked, assumes the whole beam was of a cracked section,

which is not true. To improve accuracy of a true effective stiffness of the beam for research

purposes, instead of using cracked moment of inertia, the cracked and uncracked regions of

the beam should be delineated and the moment area integrated with the respective stiffness to

find the true deflection. This, however, was not done for this comparison.

Table 6.8 Comparison of measured and calculated stiffness for the tested FRP specimens

Measured Stiffness, k/in Calculated Stiffness, k/in Specimen Kinitial* Keffective K gross transformed Kcracked

Panel A 184 88 171 87 Panel B 842 152 288 146 Panel C 593 153 511 258 Panel C2 296 217 558 283 Beam D1 235 118 279 141 Beam D2 143 102 295 150 Beam D3 167 126 296 150 Beam E1 421Γ 249Γ 291 82 Beam E3 381Γ 272Γ 229 79 Beam E5 not available --- 314 83 Beam E2 (Fatigue Beam)

410Γ 236Γ 291 82

Note: * Kinitial is based on secant line to 16 kips Γ Stiffness based on LVDT data, other values based on actuator stroke

165

Comparing the values in Table 6.8 of the simply supported panels and beams (A, B,

C, C2, D1, D2, and D3), it can be generally stated (expect for Panels A and B) the measured

stiffnesses were below the calculated values. With many variables involved it is difficult to

definitively determine a cause for these discrepancies, although there does not seem to be a

systematic pattern of error. On the other hand, the negative moment beams performed very

well when measured stiffnesses are compared to the calculated stiffnesses. The performance

of the negative moment beams gives some confidence that a continuous FRP concrete deck

on a slab/girder system should easily meet deflection requirements.

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7 ANALYTICAL MODEL TO DETERMINE NEGATIVE MOMENT DISTRIBUTION WIDTH

7.1 Introduction

One of the stated objectives of the research program was determining a negative

distribution width over the girders for the load that will cause concrete tension cracking on

the top surface of the concrete deck. Cracking of the concrete over the girders can allow

water ingression into the deck. Water in the section, exposed to freeze-thaw cycles, will

affect the long-term durability of the bridge deck by forcing the concrete to spall and crack.

In design, to prevent concrete cracking above the girder, a distribution width of the deck

must be assumed as resisting the wheel load and the moment in that width must remain

below the cracking value. Determining the negative moment distribution width employs the

same procedure as used for the positive moment distribution, except it is defined over the

girder instead of in the middle of the span.

Another difference between the positive and negative moment distribution widths is

that the negative moment distribution width was determined by analytical modeling with

shell finite elements modeling the 2-D deck. The experimental results were used, however,

as a benchmark of verification for the analytical model. The instantaneous elastic defections

measured for Panel C2 were used as a basis, because it’s reinforcement system (added

aggregate coverage on the surface of the FRP deck form) is closest to representing the

reinforcement system to be used in the prototype bridge.

The purpose of this Chapter was to layout the methodology taken to derive the

negative moment distribution width. First part of this chapter will provide an overall

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description of the process. Another part of the methodology will define the section

properties of the orthrotropic plate model, as well as, to explain the analytical computer

program and boundary conditions used. Then the modeling methodology for the prototype

bridge will be described and finally, the distribution width at the girder will be determined

using the same analytical procedures used for the positive moment distribution width in

Chapter 6. In addition to the negative moment distribution width, the positive moment

distribution width was re-evaluated using the same model. The positive distribution width

was investigated with the wheel loads in adjacent spans that should cause the cracking

moment of the concrete on the bottom surface of the deck form.

7.2 Overall Description of the Methodology

First, the cross-sectional properties of the orthogonal FRP reinforcement system were

calculated in each direction for the positive moment region. These values were used to

define shell element properties in the structural analysis computer program SAP 2000 for the

concrete deck panel test of Panel C2. The experimental Panel C2 test was duplicated

analytically in SAP 2000. The instantaneous deflection values and profiles in the elastic

region of the analytical model were compared to the experimental values. Once the quality

of the analytical model was verified experimentally (by comparing deflections) for the

positive moment regions, the same approach was now used for a model of the prototype

bridge deck with wheel load centered in adjacent spans to determine the negative moments

developed over the girder. These moments were compared to the cracking moment derived

in Section 6.4.1 from the negative moment beam tests. Then the wheel load values in the

model were adjusted until the loads created a negative moment over the girder that would

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just cause the concrete to crack in tension. At this load, the moment profile of the section

over the girder was used to determine the distribution width. This is the distribution width

that can be used in design to examine to see if certain wheel loading on this FRP

reinforcement system will cause cracking.

The objective was to determine the distribution width of the negative moment region

just as the concrete cracking occurs. With that stated, and reasoning that a continuous

section of the model will be stiffer than a simply supported concrete deck panel, an initial

assumption was made that the entire section should remain with gross section of inertia (and

full composite action) up to the first concrete crack over the girder. All section properties

were defined, however, so that each individual shell element’s cross sectional properties

could be easily changed in the model to it’s cracked section if the positive cracking moment

in that shell element was reached before the negative cracking moment developed. The

model was run to prove the initial assumption on stiffness of a continuous section and the

model was validated by the experimental results.

7.3 Defining the Section Properties

The first step was to define positive moment transformed gross and cracked section

properties for the cross section of the orthrotropic FRP reinforced deck in both directions, see

Appendix D for the illustrated parameters used for a typical 18” cross section. The

“longitudinal” direction is orientated perpendicular to the girder supports. The transverse

direction was the direction of the temperature and shrinkage reinforcement of the top bi-

directional grid, and was parallel to the supports. The cracked section properties of the FRP

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reinforced system were calculated according to Equation 7.1 which was provided by ACI

Guide 440H.

Icr = (bd3/3)k3 + nfAfd2 * (1-k)2 (7.1)

Where,

k = (2ρfnf + (ρfnf)2)1/2 - ρfnf (7.2)

The variables are defined as,

b = width of retangular cross section, in; d = effective reinforcement depth, in; nf = modular ratio of FRP to concrete; Af = area of FRP reinforcement, in2; and ρf = FRP reinforcement ratio. The gross section property was transformed to include the FRP deck form, which is slightly

less than a true gross sectional value of a true rectangular section (see Appendix D). The

difference between the gross and the gross-transformed section is due to the voids in the

concrete created by the corrugations of the FRP deck form and the fact that the modular ratio

between the two materials is slightly above unity.

To accurately model instantaneous deflection in the experimental test of Panel C2,

four possible moment of inertias for each shell element were defined:

1. Longitudinal transformed gross moment of inertia; 2. Longitudinal cracked moment of inertia; 3. Transverse transformed gross moment of inertia; and 4. Transverse cracked moment of inertia.

Incidentally, using Equation 7.1 to calculate the transverse cracked moment of inertia,

(number 4 in the above list) the area FRP material in the two corrugations was conservatively

ignored and the area, Af, of bottom material of the FRP deck form alone was used, however,

this moment of inertia was never used in this model analysis. After establishing the moments

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of inertia, the respective cracking moments were calculated according to Equations 6.6 and

6.7. All of the aforementioned moment of inertia and cracking moment values are shown in

Appendix D.

7.4 Finite Element Modeling

To model the orthogonal properties for Panel C2, and to model the continuous bridge

deck, the structural analysis computer program SAP2000 Nonlinear Version 7.44 was used.

The model of Panel C2, as well as the bridge deck model, was defined with shell sections.

For the bridge model, it was assumed that the elastic deflections were small enough not to

invoke any involvement of membrane forces.

7.4.1 Verifying Experimental Results, Modeling Panel C2

The concrete deck panel was modeled with 6” x 6” square shell elements. Each

element was initially set to the gross-transformed section properties in each direction of the

orthogonal material. The supports were model as simply supported with pin supports along

bearing line and rollers on the other bearing line. A load was applied as a pressure to an area

representing the wheel load as illustrated in Figure 4.1. Three load cases were created:

deadload only, liveload (wheel load) only, and a combination thereof. The combination load

case was analyzed to examine for any uplift forces at the supports. Otherwise all

experimental data comparisons from testing excluded the deadload, since the instruments

were applied after the panels were in the testing setup (preventing the ability to measure

deadload deflections and strains).

The first objective was to verify the analytic model accuracy by comparison with the

experimental liveload deflections in the elastic range. Therefore, the first step for

171

verification was to simulate liveload application and to compare elastic analytical deflections

at a common point with the experimental deflection measurements. So, at 8 kip intervals the

analytical deflection at 24” from the centroid of the panel along the center of span was

compare to the experimental results at the same location and load. At the onset, all cross-

sectional properties had a gross-transformed moment of inertia. In addition, the reactions at

each load increment were monitored for any indication of tension forces. If tension was

encountered in the analytical model, the support at a node was released to prevent the

introduction of hold down forces due to an improper support condition model.

Figure 7.1 illustrates the comparison of liveload deflections, at 8 kip intervals

between the model for Panel C2 and the respective inelastic experimental test results. The

graph clearly displays a close relationship between the measured elastic deflections from the

test of Panel C2 and the analytical model. The analytical model did not require any reduction

of section properties from the initial gross-transformed moment of inertia. The comparison

was conducted up to 48 kips because the loading for the investigation of the negative

moment distribution was limited to the elastic region just before cracking over the girder

occurs, which was below 48 kips. According to the model results, the longitudinal positive

cracking moment was not reached, directly under the wheel load, at a liveload of 16 kips

(M16kip = 4.80 and Mcr = 5.50 kip-inch per inch width) Incidentally, examining the

combination load case revealed that each of the four corner nodes did experience net uplift

forces at the 24 kip load interval and thus were released, but no other nodes showed uplift

forces up to 48 kip.

172

0

10

20

30

40

50

60

0 0.01 0.02 0.03 0.04 0.05Deflection, in

Load

, kip

s

Panel C2 Model Uncracked Figure 7.1 Comparison of experimental (inelastic test) and analytical deflections at 24” from

center of load along center span

With the encouraging sign that experimental, data from inelastic test, and the

analytical measurements were very close at one location, the next step was to compare

deflection profiles. Comparing deflection profiles give an indication of lateral stiffness,

which was critical for building a valid analytical model to investigate load distribution.

Therefore, the deflection profiles were compared using the measurements from the elastic

test of Panel C2 along two lines, at center span and A24 or 24” toward the support from the

center span (see Figure 4.8). The elastic test deflections were used because more ±1.0”

LVDTs were set up along the center span, thus providing a better deflection profile. The

profile measurements from the elastic test and the model, at a 16 kip load, are compared in

Figures 7.2 and 7.3. The results in Figure 7.1 were based on a span of 8’-0”. With the

magnified deflection scale of Figure 7.2 it was clear that deflections of the test panel were

not exactly matched by the analytic model. A second analysis was undertaken with a revised

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effective span length. Analytical results are sown in Figures 7.2 and 7.3 with an effective

span of 8’-0” center-to-center of bearing, and with 7’-4” clear span between front of

bearings.

-0.020

-0.015

-0.010

-0.005

0.000-54 -36 -18 0 18 36 54

Distance from the Center, in

Def

lect

ion

from

16

kip,

in.

C2 Elastic Model 7 ft -4 in span Model, 8 ft span

Figure 7.2 Deflection profile for experimental (elastic test) and analytical measurements at

center span

-0.012

-0.010

-0.008

-0.006

-0.004

-0.002

0.000

-54 -36 -18 0 18 36 54

Distance from the Center, in.

Def

lect

ion

fro

m 1

6 ki

p, i

n.

C2 Elastic Model 7 ft -4 in span Model, 8' span

Figure 7.3 Deflection profile for experimental (elastic test) and analytical measurements along A24

The deflection profiles are quite similar in shape and it was encouraging that the

analytical gross-transformed section properties defined for the model were validated by the

experimental results. It is interesting to note that the experimental deflection values were

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approximately halfway between the analytical values obtained using center-to-center and

clear spans, which makes sense of a true span being from front quarter bearing to front

quarter bearing, or 7’-8”. At this point the conclusion was made that the positive moment

region can be accurately modeled. In Section 6.4.1, the negative moment region section

behavior was validated by good agreement between the loads that caused concrete cracking

over the support (Beams E1 and E3) and the analytical calculations using the gross-

transformed section properties. With analytical validation for modeling of both the negative

and positive moment regions the next step was to model the prototype bridge deck.

7.4.2 Modeling the Prototype Bridge Deck

The bridge deck model had the same geometric measurements that will be seen in the

Wisconsin DOT prototype bridge described in Appendix A. The center-to-center girder

spacing is 8’-8” and the bearings on top of each girder are 16” wide. Because of the presence

of shear stirrups from the top of the girders cast into the concrete deck, there will be some

flexural continuity between the deck and the girder. In this model that flexural restraint was

not directly modeled and pinned connections were placed along each edge of the bearings.

Thus the model span was taken as the clear span and the flexural restrain of the girders was

indirectly accounted for by having more than one node, in the span direction, pin connected

to the girder/support. Although a deck length should theoretically be considered equal to the

bridge length, a conservative length of about twice the center-center span length, 16’, was

chosen for the model. Wheel loads were placed in the center of two adjacent spans forcing

an assumed maximum moment over the top of the center girder. Spacing of wheel loads in

accordance with AASHTO could have be considered here, but it was conservatively assumed

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that at some point in the life of the concrete deck the undesirable condition of two loads

being placed centered in adjacent spans would occur. Two additional spans were added to

the model, one to each end of the loaded spans. A 28-day concrete strength of 4000 psi,

specified by the WisDOT for bridge use, was assumed.

Learning from the process of modeling Panel C2 elastic deflections, the section

properties remained equal to the gross-transformed moment of inertia. The moment of

inertia properties would not change if the negative moment section cracks first. With the

positive and negative cracking moment essentially the same, Mcr (Mcr = 4.79 kip-in / in

width, f’c = 4000 psi), the negative moment over the girder should consistently be larger than

the positive moment in the span as the load increases in the elastic region and, thus, crack

first.

Essentially the process of modeling involved meshing the concrete deck into 6”

square shell elements, placing the pinned connections and loading the center of the adjacent

spans until the maximum moment over the girder reached the cracking moment. At this load,

the magnitude was noted as the load to cause longitudinal flexural cracking (in direction of

traffic and girder centerline) of the concrete over the girder. Then the moment profile along

the cross-section, parallel to the girder centerline, was subjected to Equation 7.3.

= (2 * m(x)dx) / mDistribution Width0

96"

maximum (7.3)

Equation 7.3 was another approach to determine the distribution width. It is

fundamentally the same as Equation 6.2, and described in Section 6.2.1, but the method to

determine the total moment (the numerator) is different. Since there was a 16” bearing span

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acting as the top of girder, it is more accurate to use the analytical model moment profile to

determine the total moment than using a 2-D beam formula (Equation 6.2 determined the

total moment with a simple 2-d beam formula, P*L/4). Equation 7.3 determined the negative

moment distribution width by summing the area under the moment distribution profile, the

total moment, and divided it by the cracking moment, Mcr, or the maximum moment per inch

of width (as will be explained later, this same method was used to analytically determine the

positive moment distribution width in the middle of the span). Figure 7.4 illustrates the

negative moment distribution profile from the analytical model for the section above the

front edge of the bearing on the center girder where the maximum negative moment

developed for liveload of 35 kip wheel load (selfweight neglected) centered in adjacent

spans.

-7

-5

-3

-1

1

3

5

-96 -80 -64 -48 -32 -16 0 16 32 48 64 80 96

Distance from center of load, inches

Mom

ent,

kip-

in/ i

nch

wid

th

Negative Moment Positive Moment

+Mcr or Mmax =4.79 k/in per inch width

-Mcr or Mmax =4.79 k/in per inch width

Figure 7.4 Moment distribution profiles over a 16’ width

The area under (actually above) negative moment distribution curve and the total

moment was 347 inch-kips at a Mcr or Mmax of 4.79 inch-kip per inch of width, which

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occurred at a liveload of 35 kips centered in adjacent spans. Using Equation 7.3, the negative

moment distribution width at the onset of longitudinal flexural concrete cracking over the

girder was 73”. Figure 7.5 is an illustration of a three dimensional moment profile in the

bridge model.

Figure 7.5 Negative moment profile of prototype deck for negative moment

The positive moment distribution was also determined from the same model. Wheel

loads in the adjacent spans were adjusted until the longitudinal positive cracking moment

(cracking in the directions of traffic) was reached in the center of the adjacent spans. The

magnitude of the load was noted and moment distribution profile along the center span (in

the direction of traffic) was also graphed in Figure 7.4. The area under positive moment

distribution curve (the total moment) was 271 inch-kips at a Mcr or Mmax of 4.79 inch-kip per

inch of width, which occurred at a liveload of 28 kips centered in adjacent spans. Again,

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using equation 7.3, the moment distribution width at the onset of the concrete cracking at the

bottom of the concrete deck at mid-span was 57.5”. This brings about one final assumption

for the negative moment distribution width, that the cracking in the positive moment region,

which occurs at 28 kip wheel loads centered in adjacent spans, does not affect the negative

distribution width at 35 kip. Another words, with wheel load centered in adjacent spans,

longitudinal concrete cracking in the positive moment region (bottom of the slab) will occur

before longitudinal concrete cracking in the negative moment (top of slab and over the

girder)

The positive and negative moment distribution widths derived from the model of the

prototype bridge cannot be directly compared to AASHTO distribution width, equation 6.3,

used for the experimental concrete deck panels, because each represents a different support

condition. According to AASHTO-LFD Bridge Design Specification, 16th edition, the

experimental concrete deck panels represent support conditions of a slab supported on

abutment or piers, AASHTO 3.24.3.2, Case B. The prototype bridge deck model represents a

slab supported on beams or stringers, AASTHO 3.24.3.1, Case A. The design liveload

moment for Case A, given by Equation 7.4, and the total moments for the negative and

positive distribution curves derived by the analytical model, however, can be relatively

compared on the basis of moment per foot width. It should be noted that Case A is

presumably for strength design and the positive and negative moment distribution widths for

a slab-stringer support condition are serviceability issues, so strictly speaking, this was not

comparing “apples to apples”.

LLM = P(S+2)/32 ft-lb (7.4)

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For an HS20-44 truck wheel load without impact on the prototype bridge, P is 16,000

pounds and the span, S, is 8.67 feet. For the given wheel load and span the design liveload

moment for the prototype bridge deck, LLM, would be 64 kip-inches per foot width (5.34

kip-inch per inch width). This design moment can be reduced by 20% if the slab is

continuous over three or more stringers, which is the case for the prototype bridge.

Therefore, the design moment is 51.2 kip-inches per foot width (4.26 kip-in per inch width).

The LLM to cause longitudinal cracking for the negative and positive moment distribution

curves were derived by determining the area ±6” each side of Mmax under the curves (see

Figure 7.4). The LLM per foot was calculated as 56.5 and 54 kip-inches per foot width (4.71

and 4.50 kip-inch per inch width) for negative and positive moment to cause longitudinal

cracking. Conservatively, the LLMs to cause longitudinal cracking over the girder and in the

slab are greater than the LLM for the service design moment prescribed by AASHTO.

It should be noted that selfweight of the deck was not included. For the sake of the

real prototype bridge, the selfweight of the deck will contribute 11% and 6% of the liveload

moments produced by the wheel loads used for the negative and positive moment distribution

widths, respectively.

180

8 CONCLUSIONS

8.1 Introduction

In the introduction of this thesis solid reasons were presented for the use of FRP

products to reinforce concrete bridge decks to increase long-term durability. FRP as a

material cannot replace the conventional steel reinforcement, however, without proper

verification of its performance. Condensing the objectives, the purpose of this experimental

study of the FRP reinforced bridge deck was to evaluate its performance under bridge

liveloads. In the end, conclusions from this investigation will impact decisions on whether or

not the proposed FRP reinforcement scheme is adequate for bridge use.

In a bit more detail, two general objectives helped drive the experimental study of the

FRP reinforced concrete bridge decks. First, the proposed FRP reinforcement system uses an

unconventional shape and a stay-in-place deck form for the top and bottom reinforcement

layers, respectively. The unconventional bottom reinforcement scheme introduces a shiplap

joint, a structural discontinuity, and neglects the use of distribution reinforcement in the

bottom layer. These progressive issues needed to be investigated for their impact on the

reinforced deck’s ability to carry bridge wheel loads. Second, the absence of a specific

design guide for this type of FRP reinforcement in concrete bridge decks leaves an engineer

to use an “ad-hoc” collection of design guides. The engineer is without the typical design

tools and equations used for steel reinforced bridge decks. Therefore, determining the mode

of failure and a distribution width are a couple of empirical objectives that will help simplify

the design process and build confidence that the studied AASHTO equations used in this

181

thesis to design a steel reinforced concrete bridge deck can be used for this specific FRP

reinforcement system.

8.2 Observations and Conclusions

Examining the results, many conclusions and observations could be and have been

drawn. In this section, only the major observations and conclusions, however, will be

presented. In addition, the qualitative and quantitative performance observations and

conclusions below will, in general, relate back to the stated objectives presented in the

introduction and are presented in bulleted form. Finally, the following observations and

conclusions will be separated into categories related to serviceability, strength, detailing and

construction issues.

8.2.1 Serviceability

1. From the test results from this thesis and according to the conclusions of Sanders

(2001), the added aggregate coverage alone increased the resistance to longitudinal

concrete cracking above the shiplap joint without distribution reinforcement, and

thus significantly increased the load required, from 5 kip (Panel C) to 47 kip (Panel

C2), to initiate the opening of the shiplap joint directly under the wheel load.

2. An analytical model for the prototype bridge, verified by experimental results in the

elastic region of Panel C2, indicated that the negative and positive moment caused

by wheel loads centered in adjacent spans will be distributed laterally 73” over the

girder and 57.5” at midspan, respectively. The analytical model also indicates the

loads centered in adjacent spans to cause a negative moment crack over the girder is

about 31 kip, including selfweight of the deck.

182

3. The amount of aggregate coverage impacts the performance of the FRP deck form

as the positive moment reinforcement. Positive moment beam tests indicated that

the partial coverage of aggregate on the top surface of the FRP deck form causes

inefficient partial composite behavior in bending.

4. Results from experimental testing of the 9’ wide concrete deck panels

conservatively indicate that the AASHTO distribution equation (AASHTO 3.24.3.2,

Case B) for a steel reinforced concrete slab supported on an abutment or piers is

applicable for use in design of the proposed FRP reinforcement system, with or

without total coverage of aggregate on the top surface of the FRP deck form. In

addition, from analytical model, the AASHTO service design equation for an HS20-

44 truck, Equation 7.4, for a steel reinforced concrete deck supported on three or

more beams or stringers is applicable for use in the prototype bridge to ensure the

prevention of longitudinal flexural cracking in the positive and negative moment

regions.

5. After 2 million cycles of load, the accelerated fatigue beam test indicated no

concern for fatigue failure of the FRP material, the mechanical bond between the

aggregate on the surface of the FRP deck form and concrete and, finally, there

seems to be some sign of stiffness loss, but not significant, for loads above the

service wheel loading. Therefore, similarly to most steel reinforced concrete bridge

decks, fatigue is not a crucial design issue.

183

8.2.2 Strength

1. Because no panels or beams failed by FRP rupture, there was a significant post-

ultimate ductility. This is a behavioral trait that can be expected from a traditional

steel reinforced concrete bridge deck.

2. The factors of safety for the ultimate strength (due to punching shear) to service

load for an HS20-44 load, without impact, are between 5 and 6 for deck forms with

partial aggregate coverage and 8.1+ for total aggregate coverage. Strength results

of Deck Panels C and C2, normalized for concrete strength, indicate that complete

aggregate coverage increased the ultimate strength and stiffness by 49% each. The

above conclusions are based on the assumption that the 4” wide FRP Fibergrate

Molded Grating as FRP reinforcement over the shiplap joints doesn’t contribute to

longitudinal strength or stiffness. Therefore, full coverage of aggregate on the FRP

deck form surfaces is recommended.

3. Concrete deck panel tests indicated that the mode of failure is punching shear and

the ACI 318 guide, Equation 6.4, can be used with a modification. If the

conventional definition of the effective depth parameter, d, is used, then a reduction

coefficient, Ag, should be used. This value of this coefficient depends upon the

amount of aggregate coverage on the top surface of the FRP deck form. With

partial aggregate coverage (original product from CDS, Inc.), Ag = 0.60, and with

aggregate to all horizontal surfaces, Ag is unity. Therefore, equation 6.4 should be

used as,

Vc = Ag * 4 * (f’c)1/2 * bo * d (8.1)

184

8.2.3 Detailing and Construction

1. From the negative moment beam tests, beam failures were based on shear followed

by a de-bonding of the form at the edge of the plaster bearing on the middle

support. Thus, the capacity of the negative moment reinforcement, the top bi-

directional grid, was never reached. It is surmised that if the FRP deck form had

been supported on the plaster bearing, the capacity of the beam would have

increased. Therefore, for the sake of providing capacity against pullout,

embedment length for positive moment reinforcement and also to properly transfer

shear to the support, the FRP deck form end must be supported upon a bearing, such

as on top of a girder or on grout.

2. From practical experience of assembling the FRP reinforcement to construct the

specimens, the system could be easily and quickly build. It was also noted that

concrete did consolidate well, using a pencil vibrator, except for areas where the top

flange bi-directional grid was positioned close to the form. For practical use,

however, this consolidation problem in a slab/girder bridge system will not be a

common experience.

Neglecting possible environmental degradation issues, this FRP reinforcement system

is applicable for use in the WisDOT prototype bridge. This conclusion is only based upon

the results from the experimental and analytical tools used in this thesis.

185

9 SUMMARY AND RECOMMENDATIONS

9.1 Design Criteria

Most Load Factor Design AASHTO criteria governing the design of a reinforced

concrete deck for a slab-on-stringer system are given in AASHTO section 3.24. A typical

continuous steel reinforced concrete slab over girders has the main steel reinforcement

oriented perpendicular to traffic. So, how does the results of the research relate to the

traditional design methodology used by AASHTO? Table 9.1 was constructed to summarize

and compare the design methodology for a traditional steel reinforced concrete deck and one

based on the results of the research on the prototype bridge conducted for this thesis. This

table is intended as a design aid for an engineer to design a deck reinforced with the proposed

FRP system. Other criteria considered in a concrete deck design, but not in the table, are

deck reinforcement details, such as minimum cover and spacing of bars, which are checked

against governing rules and must be applied with engineering judgment for the FRP

reinforcement.

By and large, designing a steel reinforced deck is a relatively easy and quick process,

however, this design methodology is based upon a long and well-documented history of

performance. When the material to reinforce the concrete is changed from steel to FRP, with

unconventional issues pointed out earlier, this design process was not certain and hence there

was a need to conduct the research for this thesis.

186

Table 9.1 Design Methodologies for Steel and FRP Reinforcement for Concrete Bridge Deck Based on 8’-8” Span

Criteria With Steel Reinforcement Based on AASHTO LFD 3.24.3.1, Case A

With FRP Reinforcement Based

on Thesis Results

FRP Reinforcement Research Conclusion

Distribution Width, in

32/(S+2) = 36” +M = 57.5” -M = 73”

AASHTO Equation is applicable for FRP, see Chapter 8.2.1, #2

Positive Moment

Design bottom layer steel reinforcement perpendicular to traffic based on MLL = P*(S+2)/32, plus impact and deadload,

Same moment as steel but check capacity using ACI 440H, where reinforcing area provided by form

FRP deck form provides over designed capacity for typical effective spans

Negative Moment

Design top layer steel reinforcement perpendicular to traffic based on P*(S+2)/32

Size “I” bar area of bi-directional grid and/or spacing based on ACI 440H

Finite Element Model shows deck remains in gross section under normal wheel loads

Distribution Reinforcement

Compute steel in bottom slab based on a percentage of main reinforcement = 67% ≥ 220/S1/2

None required Deck performed acceptably without distribution reinforcing. See 8.2.1, #1

Temperature and Shrinkage Steel

Perpendicular to main reinforcement AASHTO 8.20 minimum of 0.125 in2 per foot in each direction

Same as steel, select size and spacing of bars in bi-directional grid panel based on modular ratio

Shear Considered safe for shear

Strength capacity controlled by punching shear

Considered safe for shear

Fatigue Considered safe Performed well under 2 million cycles of load

No concern for FRP fatigue resistance, see 8.2.1, #5

9.2 Strength of FRP-SIP Reinforced Deck

From the experimental testing, the FRP system (designed according to ACI 440H for

flexural capacity) has provided satisfactory punching shear capacity under a HS44-20 truck

load, which is the generally accepted mode of failure for normal spans. Even though the

187

flexural capacity was not reached, the punching shear capacities from the concrete panel test

provided strength factors of safety of 5 to 8 times service load. The punching shear behavior

under closely spaced wheel loads, where critical shear perimeters overlap has not been

investigated but the design using ACI punching shear criteria should still apply.

9.3 Design for Serviceability

The FRP reinforcement system can provide adequate fatigue resistance when exposed

to 2 million cycles, where each load cycle induced 14% of the ultimate stress in the bi-

directional grid. Fatigue resistance under higher stress levels and exposed to reversal

stresses, however, are not known. Experimental testing also determined that the lateral load

distribution widths in the positive moment region for various span lengths were higher than

what current AASHTO criteria for steel reinforced deck allows. Use of current AASHTO

methods is acceptable. Deflection criteria were not formally addressed and still remain

uncertain for the FRP system. However, Harik et al. (1999) have determined that the SIP

FRP deck form studied in this thesis can satisfy the deflection criteria. Ohio Department of

Transportation (ODOT) was confident enough to implement this FRP deck form into an

actual bridge.

Since deflection criteria was not completely examined, it would be instructive to

compare the deflection and shear results of the positive moment beams to an actual

performance specification used for the Ohio Department of Transportation’s Salem Avenue

Bridge, where ODOT implemented the SIP FRP deck form. Table 9.2 lists the performance

of the three simply supported positive moment beams to deflection performance criteria set

188

by ODOT for the FRP deck based on reinforced concrete deck analysis. It can be seen that

the simply supported positive moment beams performed much better than the criteria.

Table 9.2 Performance of Positive Moment Beams to ODOT criteria for 3’ wide, 8” thick simply supported beams

Positive Moment Beams ODOT Criteria Span, c/c beams (ft) 8’-8” 8’-0” 9’-0” Clear Span (ft) 8’-0” 7’-6” 8’-6” Service Load (kip) 16 12 Simple Span ∆ (in) 0.068 0.112 0.96 0.125 0.18 From the analytical testing, it was shown that for an 8” concrete deck with 8’-8”

center-to-center span with section properties of the FRP reinforced concrete, that the wheel

loads to initiate longitudinal flexure cracking (in the direction of traffic) in the positive and

negative moment regions are much larger than the 16 kip service wheel load. In fact, the live

load moment derived from Equation 7.4 is less than the moment required to initiate cracking

in either region. Other span lengths and deck thickness, however, would need to be looked at

on a case-by-case basis, since no history of performance has been established. This research

is applicable for normal girder spacing between 8’ and 12’ and any spacing outside the

ranges of the concrete deck panel tests may deviate from the results described here.

9.4 Further Investigation

Even though this FRP system has performed well relative to serviceability and

strength criteria, there are some issues that are still unknown and should be investigated

further. The minimum edge bearing of the FRP deck form to ensure that the form does not

pull out and delaminate before the punching or flexural shear capacity is reached is not

certain. The test specimens performed adequately with little or no bearing, but improved

behavior could be achieved. Another outstanding issue deals with reversal of and higher

189

levels of stress effects upon the fatigue capacity. Long-term environmental effects upon the

FRP deck form could pose a durability problem and should be considered for further study.

Other tests (Helmueller, 2001) have shown that moisture collection between the FRP does

not lead to serious de-bonding problems when subjected to freeze thaw action and the FRP is

not susceptible to corrosion. It may be beneficial to further investigate the possibility of

longitudinal concrete cracking above the shiplap joints between the FRP deck forms. As a

final point, other issues related to detailing and construction, such as tying down the FRP

deck forms prior to concrete placement, need to be considered, but these construction

solutions may be best developed through ingenuity and creativity in the field.

9.5 Improving the FRP Reinforcing

To conclude the summary of this thesis, improvements to the FRP reinforcement

system are suggested based on test results, experienced gained from the construction of

specimens, and some insight of future construction issues. Most of the improvements

reflected in the test results are related to the SIP FRP deck form. First, the deck form should

incorporate complete coverage of aggregate to all horizontal surfaces to ensure efficient bond

and thus increased performance in strength and stiffness. In addition, it has been shown that

through the lack of structural continuity, the shiplap joints were a point of weakness for

strength. This could be improved in two ways: 1.) the shiplap joint could be redesigned to

provide a shear connection in both directions between adjacent deck forms, and 2.) the deck

forms could fabricated at larger width increments, e.g., 3’, and thus reduce the number of

joints (however this depends on fabrication, shipping and handling limitations). It also may

be economically beneficial to investigate the redesign of the FRP deck form using less

190

material. Adding aggregate and reducing the cross-sectional area of the FRP material could

structurally economize the flexural capacity of the system and reduce cost of the product.

9.6 Constructability

This new system should substantially reduce deck construction time because of the

use of lightweight prefabrication components that can be placed quickly without heavy

equipment. From a constructability perspective, there might be an easier way to provide

temperature and shrinkage reinforcement continuity between adjacent bi-directional grid

panels. In the research, a smaller 12” wide splice grid was tied below the main

reinforcement, but from experience this could require too much time to construct properly. A

natural solution for a composite girder/slab superstructure system that requires negative

moment reinforcement over a pier or bent is to run a percentage of this reinforcement to the

abutments. In the process, the continuation of reinforcement will also provide capacity

against possible concrete cracking above the shiplap joints and act, in some respect, as

distribution reinforcement.

191

REFERENCES

Alampalli, S., O’Connor, J., and Yannotti, A. P. (2000), “Fiber Reinforced Polymer Composites for Superstructure of a Short-Span Rural Bridge.” Albany, N.Y.: Transportation Research and Development Bureau, New York State Department of Transportation, (Special report; 134) Alkhrdaji, T., Nanni, A., and Mayo, R. (2000). “Upgrading Missouri Transportation Infrastructure: Solid RC Decks Strengthened with FRP Systems.” 79th Annual TRB Meeting (CD-ROM), National Academy of Science, Washington, D.C. Bank, L.C., Xi, Z. and Munley, E. (1992). “Tests of Full-sized Pultruded FRP Grating Reinforced Concrete Bridge Decks.” In Materials: Performance and Prevention of Deficiencies and Failures, (ed. T.D. White), ASCE Materials Engineering Congress, Atlanta, GA., August 10-12, pp. 618-631. Bank, L.C., and Xi, Z. (1993). “Pultruded FRP Grating Reinforced Concrete Slabs.” In Fiber-Reinforced-Plastic for Concrete Structure – International Symposium, (eds. A. Nanni and C.W. Dolan), SP-138 American Concrete Institute (ACI) pp. 561-583. Bank, L.C., and Xi, Z. (1995). “Punching Shear Behavior of Pultruded FRP Grating Reinforced Concrete Slabs.” Proc. 2nd Int. Symp. On Non-Metallic (FRP) Reinforced for Concrete Structures, (ed. L. Taerwe), E & FN Spon, London pp. 360-367. Bettigole, Neal H. and Robinson, R. (1997). Bridge Decks: Design, Construction, Rehabilitation, Replacement. American Society of Civil Engineers (ASCE) Press, New York, 1997.

Bradberry, T.E. (2001). “FRP-Bar-Reinforced Concrete Bridge Decks.” 80th Annual TRB Meeting (CD-ROM), National Academy of Science, Washington, D.C. Chajes, M., Shenton, H., and Finch, W. (2001). “Performance of a GFRP Deck on Steel Girder Bridge [Paper 01-3290].” 80th Annual TRB Meeting (CD-ROM), National Academy of Science, Washington, D.C. Dietsche, J. S. (2002). “Development of Material Specification for FRP Structural Elements for the Reinforcing of a Concrete Bridge Deck.” MSCE Thesis, Department of Civil and Environmental Engineering, University of Wisconsin-Madison. Harik, I., Alagusundaramoorthy, P., Siddiqui, R., Lopez-Anito, R., Morton, S., Dutta, P. and Shahrooz, B. (1999). “Testing of Concrete/FRP Composite Deck Panels.” Proceedings of the Fifth ASCE Materials and Engineering Congress (ed. L. C. Bank), Cincinnati, May 1999, ASCE, Reston, VA. pp. 351-358.

192

Helmueller, E.J. (2001). “The Effect of Freeze-Thaw and Aggregate Coating on the Bond Between FRP Stay-in-Place Deck Forms and Concrete.” Independent Study, Department of Civil and Environmental Engineering, University of Wisconsin-Madison. Kumar, S.V., and GangaRao, H.V.S. (1998). “Fatigue Response of Concrete Decks Reinforced with FRP Rebars.” Journal of Structural Engineering, 124(1), ASCE, 11-16. Lopez-Anido, R., Howdyshell, and P., Stevenson, L. D. (1998). “Durability of Moduar FRP Composite Bridge Decks Under Cyclic Loading.” 1998 Durability of Fibre Reinforced Polymer (FRP) Composites for Construction (CDCC). pp. 611-622. University of Sherbrooke, Quebec, Canada. Matthys, S., and Taerwe, L. (2000). “Concrete Slabs Reinforced with FRP Grids. I: One-Way Bending.” Journal of Composites for Construction, ASCE, 4(3), 145-153. Matthys, S., and Taerwe, L. (2000). “Concrete Slabs Reinforced with FRP Grids. II: Punching Resistance.” Journal of Composites for Construction, ASCE, 4(3), 154-161. Priestley, M.J.N., “The US-PRESSS Program Progress Report.” 3rd Meeting of the U.S.-Japan Joint Technical Coordinating Committee on Precast Seismic Structural Systems (JTCC_PRESSS), San Diego, CA, November 18-20, 1992 Reising, R.M.W., Shahrooz, B.M., Hunt, V.J., Lenett, M.S., Christopher, S., Neumann, A.R., Helmicki, A.J., Miller, R.A., Kondury, S., Morton, S. (2001). “Performance of a Five-Span Steel Bridge with Fiber Reinforced Polymer Composite Deck Panels [Paper 01-0337].” 80th Annual TRB Meeting. Sanders, B.M. (2001). “Structural and Physical Tests of CDS-FRP Reinforced Deck Form and Concrete.” Independent Study, Department of Civil and Environmental Engineering, University of Wisconsin-Madison. Taly, N. Design of Modern Highway Bridges. McGraw-Hill, New York, 1998. AASHTO (1996). Standard Specifications for Highway Bridges. 16th Ed., Washington, D.C. AASHTO (1998). LRFD Highway Bridge Design Specifications. 2nd Ed., Washington, D.C. AASHTO (2000). Manual for Condition Evaluation of Bridges. 2nd Ed., Washington, D.C ACI. (1999). Building Code Requirements for Structural Concrete and Commentary. ACI 318-99 and ACI 318R-99. American Concrete Institute, Farmington Hills, Michigan. ACI. (2001). Guide for the Design and Construction of Concrete Reinforced with FRP Bars. ACI 440.1R-01. American Concrete Institute, Farmington Hills, Michigan.

193

ASTM. Standard Method of Testing Compression Strength of Cylindrical Concrete Specimens (C39-93). Philadelphia: American Society for Testing and Materials, 1993. Wisconsin Department of Transportation (1996). Standard Specifications for Highway and Structure Construction. 1996 edition.

194

APPENDIX A

195

DESCRIPTION OF PROTOTYPE BRIDGE

The bridge targeted for this FRP reinforcement system is new construction on US Highway

151 (A.D.T. of 18,600) over State Highway 26 near the city of Waupun, Wisconsin. This is

a twin, 2-lane bridge structure with two continuous 32.7 m (107’) spans. The girders are 1.37

m (54”) deep pre-stressed concrete “I” beams, spaced at 2.65 m (8’- 8”). The deck is 200

mm (8”) thick (excludes wearing surface), 12.75 m (43’) wide with a total deck area of 855.6

m2 (9,200 ft2). Construction is expected to begin in the spring of 2003.

The proposed innovative construction technologies should improve upon current

construction practices. One proposed technology is simplifying the concrete reinforcement

assembling process and, therefore speeding up construction. The pultruded SIP FRP deck

form and the pultruded grid panels will be pre-fabricated and pre-sized units and delivered to

the job site. The deck forms will be placed in rapid fashion without the need for time-

intensive falsework or formwork, which is usually needed in conventional concrete deck

pours. With the companion bridge decked with a traditional steel reinforcement system, a

“field laboratory” will be created to compare in-service performance behavior. The two twin

bridges will be monitored to provide long-term data comparisons in identical harsh

environments.

196

DESIGNED FRP REINFORCEMENT FOR A TYPICAL 18” CROSS SECTION

= d


1 1/2"

1/4"

8"

2"

1 1/2"

6.93

"

1'-6"

3"

FRP Deck Form

Positive moment section

Area of FRP deck form = 11.89 in2 Area of 2” “I” bars = 2.77 in2

1/4"

1 1/2"

8"

1 1/2"

2"

1'-6"

d = 5.69"

3"FRP Bi-directional Grid

FRP Deck Form

Negative moment section

197

APPENDIX B

198

ILLUSTRATION OF STRAIN GAUGE LOCATIONS FOR THE TEST SPECIMENS

(Specimen Assembly)

2" Ø PVC

Reinforcement Clearence = 1-1/4"Round Bars = 4" o.c.

L

"I" Beams = 4" o.c.

Bi-Directional Grid Bar Spacing: C

Strain Gauges

Location to Center of Longitudinal Grid Splices

A

Location of Panel Letter on Concrete Deck Surface

FRP Panel Width

Panel A

Concrete Deck Width

Gauge Locations

FRP Deck Form A3

FRP Deck Form A2

FRP Deck Form A2

199


(Specimen Assembly)

"I" Beams = 4" o.c.

Bi-Directional Grid Bar Spacing:

Round Bars = 4" o.c.

2" Ø PVC

Reinforcement Clearence = 1-1/4"

LC

Strain Gauges

Location of Panel Letter on Concrete Deck Surface


B

FRP Panel Width

Panel B

Concrete Deck Width

Gauge Locations

FRP Deck Form B3

FRP Deck Form B2

FRP Deck Form B1

200


(Specimen Assembly)

Reinforcement Clearence = 1-5/8"Round Bars = 4" o.c.

Bi-Directional Grid Bar Spacing: L

"I" Beams = 4" o.c.

C

2" Ø PVC

Strain Gauges

Location of Panel Letter onConcrete Deck Surface


FRP Panel Width

C

Gauge Locations

Concrete Deck Width

Panel C

FRP Deck Form C3

FRP Deck Form C2

FRP Deck Form C1

201


(Specimen Assembly)

Panels E

Panel E1

Panel E1

Panel E1

E1

Location of Panel Letter onConcrete Deck Surface

E2

E3

2" Ø PVC

Round Bars = 4" o.c."I" Beams = 4" o.c.

For All Panel E's, Bi-Directional Grid Bar Spacing:

For All Panel E's, Reinforcement Clearence = 1-1/2" @ CL

Gauge Locations

202

WISCONSIN DOT CLASS D CONCRETE DESINGN MIX

(provided by M&M Concrete) DATE: 01/11/2001 ATTN: Dave PROJECT: U.W. Madison CONTRRACTOR: IF Corp. PART OF STRUCTURE: Panels CLASS OF CONCRETE: WDOT Class D MIX DESIGN NUMBER: 208

MATERIAL 610 WEIGHT VOLUME Free Moisture W/C PORTLAND CEMENT 1.0 610 3.10 3.15 0.45 GRANCEM CEMENT 0.0 0 0.00 2.83 FLYASH 0.0 0 0.00 2.65

SLUMP

COURSE AGGREGATE ¾” 1802 10.75 2.66 1.00% 4in. Max COURSE AGGREGATE 1-1/2” 0 0.00 2.66 1.00% FINE AGGREGATE 1216 7.22 2.62 3.00% WATER 275 4.41 AIR 6.00% ± 1.0 6.00 1.62 Air Mix 250 1.00 oz/cwt 6.10 WR 0.00 oz/cwt 0.00 MR 3.00 oz/cwt 18.30 Eucon 37 0.00 oz/cwt 0.00 Accelgaurd NCA 0.00 oz/cwt 0.00 CaCl 0.00 percent 0.00 Fiberstrand F 0.00 bag/yard 0.00 3903 27.10

MATERIAL TYPE SUPPLIER PORTLAND CEMENT ASTM-C150 HOLNAM INC. GRANCEM CEMENT ASTM-C989 HOLNAM INC. FLYASH ASTM-C618 C Mineral Solutions COURSE AGGREGATE #1 ASTM-C33 Prairie Avenue COURSE AGGREGATE #2 ASTM-C33 Prairie Avenue FINE AGGREGATE ASTM-C33 Mann Brothers Air Mix 250 ASTM-C260 Euclid Chemical Company WR ASTM-C494 Euclid Chemical Company MR ASTM-C494 Euclid Chemical Company Eucon 37 ASTM-C494 Euclid Chemical Company Accelgaurd NCA ASTM-C494 Euclid Chemical Company CaCl ASTM-C494 Euclid Chemical Company Fiberstrand F ASTM-C1116 Euclid Chemical Company

Prepared By: Jamie M. Kuchnickl

203

APPENDIX C

204

EXPERIMENTAL INSTRUMENT LOCATION FOR ULTIMATE TEST

PANEL B, TOP CONCRETE SURFACE

CL

Panel B Span

9' Width

(9'-10", nts)

4" Potentiometer



SG17

12"9"

LVDT15

12"

LVDT16

12"

LVDT17

Control

SG6

15"

6"

7"

SG4

LC

6"

205


PANEL B, BOTTOM FRP SURFACE

SG2 9' Width

SG7


CL

13"

Control

SG5

21"

9"

SG3


LVDT13 to monitor horiz.joint opening (transverse)

1-1/2"


(9'-10", nts) Panel B Span

6"


206

EXPERIMENTAL INSTRUMENT DATA RECORD FOR THE ULTIMATE TEST

PANEL B, 3/12/01

Instrument Type

LabView Channel


Comments

Load 1 -40.0 Stroke 2 -2.0 4” Potentiometer 3 0.615 12” from load ±0.1” LVDT13 4 0.0198 Joint opening SG7, FRP 5 0.005185 Below SG4 ±1” LVDT15 6 0.2121 24” from load

±1” LVDT16 7 0.2060 36” from load

±1” LVDT17 8 0.2047 48” from load SG2, FRP 9 0.004107 Center SG3, FRP 10 0.002718 Below SG17 SG4, Concrete 11 0.008751 Above SG7 SG5, FRP 12 0.002758 Below SG6 SG6, Concrete 13 0.002760 Above SG5 SG17, Concrete 14 0.002597 Above SG3

207


PANEL C, TOP CONCRETE SURFACE

CL 9' Width12

"

LCControl

LVDT17

LVDT16

LVDT15

12"



4" Potentiometer

11-3/4"

4-1/2"

(8'-0", nts) Panel C Span

SG178-

3/4"

SG6

SG4

24-1

/2"

19-1

/2"

15-1/4"

208


PANEL C, BOTTOM FRP SURFACE

LVDT13 to monitor horiz.joint opening (transverse)

Control

LC

SG7, SG4 denotes strain gauge LVDT14, LVDT15,... denotes LVDT

16-3

/4"

17-3

/8"

SG2LC

2"9-

1/4"SG3

SG5

9' Width


SG7

4-1/2" Panel C Span

(8'-0", nts)


209

EXPERIMENTAL INSTRUMENT DATA RECORD FOR THE ULTIMATE TEST

PANEL C, 3/26/01

Instrument Type

LabView Channel


Comments

Load 1 -40.0 Stroke 2 -2.0 4” Potentiometer 3 - 0.615 12” from load ±0.1” LVDT13 4 -0.01984 Joint opening SG7, FRP 5 -4.71 Below SG4 ±1” LVDT15 6 -0.2103 24” from load

±1” LVDT16 7 0.2073 36” from load

±1” LVDT17 8 -0.2075 48” from load SG2, FRP 9 -2.898 Center SG3, FRP 10 -2.762 Below SG17 SG6, Concrete 11 -2.282 Above SG5 SG5, FRP 12 -2.829 Below SG6 SG4, Concrete 13 -2.578 Above SG7 SG17, Concrete 14 -2.600 Above SG3

Note: SG calibration constants are x1000

210

EXPERIMENTAL INSTRUMENT LOCATION FOR THE ELASTIC AND THE ULTIMATE TEST

PANEL D3, CONCRETE TOP SURFACE

Span8'-8"


3'

Control

SG2

SG3

4" Potent.

7"

LVDT13

CL7"

4" 3"

4"

SG6

18"

16"Mechanical SG

6 1/2"

5 3/4"

PANEL D3, BOTTOM FRP SURFACE

SG7

16"

1'-5

"

CLControl

FRP deck form

FRP deck form4"

4"

Span8'-8"

3'SG4

SG5

211

EXPERIMENTAL INSTRUMENT DATA RECORD FOR THE ELASTIC AND THE ULTIMATE TEST

PANEL C, 4/17/01


Instrument Type

LabView Channel

Elastic Inelastic

Comments

Load 1 -5.0 -40.0 50k/200k Stroke 2 -0.2 -2.0 2”/10” 4” Potentiometer

3 -0.610 -0.610

±0.1” LVDT13

4 0.0245 0.0245

SG2, Concrete 9 4.2066 4.2066 Above SG4 SG3, Concrete 10 2.7123 2.7123 Above SG5 SG4, FRP 11 8.8229 8.8229 Below SG2 SG5, FRP 12 2.8206 2.8206 Below SG3 SG6, Concrete 13 2.4511 2.4511 Above SG7 SG7, FRP 14 2.6275 2.6275 Below SG6 Mechanical SG used to verify electronic SG data, 4” spacing

Note: SG calibration constants x1000

212

APPENDIX D

213

DESCRIPTION OF PRIESTLEY STANDARD METHOD TO CALCULATE A

STIFFNESS AND A YIELD STRENGTH FOR A SPECIMEN TEST RESULT

0

10

20

30

40

50

60

70

80

90

100

0 0.5 1 1.5 2 2.5 3 3.5

Displacement, in

Lo

ad, k

ips

Pmax

.75*Pmax

Disp. @ .75 Pmax Yield Disp.

Pyield

Priestley (1992) algorithm is used as a standard method to determine initial stiffness

and yield strengths for relative comparisons between specimen testing results. In general, the

sequential method to determine the relative stiffness and yield strength is defined below and

exhibited in the above illustration.

1. Determine the maximum load, Pmax, and identify the displacement result for 0.75*Pmax, “Disp. @ .75 Pmax”

2. Determine the relative stiffness, keff, by (0.75*Pmax)/(Disp. @ .75 Pmax), kip/inch

3. Determine the yield strength, Pyield, by identifying the test load which corresponds to the displacement value calculated by Pmax/keff

__________________________________________________________________________ Priestley, M.J.N., “The US-PRESSS Program Progress Report”, 3rd Meeting of the U.S.-Japan Joint Technical Coordinating Committee on Precast Seismic Structural Systems (JTCC_PRESSS), San Diego, CA, November 18-20, 1992

214

DESIGNED FRP REINFORCEMENT FOR A TYPICAL 18” CROSS SECTION

= d


1 1/2"

1/4"

8"

2"

1 1/2"

6.93

"

1'-6"

3"

FRP Deck Form

Positive moment section

Area of FRP deck form = 11.89 in2 Area of 2” “I” bars = 2.77 in2

1/4"

1 1/2"

8"

1 1/2"

2"

1'-6"

d = 5.69"

3"FRP Bi-directional Grid

FRP Deck Form

Negative moment section Panel C2 Section

Direction (f’c = 5303 psi)

Gross- Transformed Moment of Inertia, in4

Ig

Gross Moment

of Inertia, in4

Ig

+M Cracked Moment

of Inertia, in4

Icr

+M Cracking Moment Kip-in

(Equation 6.6) Mcr

-M Cracked

Moment of Inertia, in4

Icr

-M Cracking Moment Kip-in

(Equation 6.6) Mcr

Longitudinal 720 768 365 99 71 99 Transverse 623 768 284 85 n/a n/a

Section Properties for a Typical 18” wide FRP Reinforcement Section with a f’c = 5303 psi

David Dieter MS Thesis 2002

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Transcript of David Dieter MS Thesis 2002