Cyclic Seismic Testing of Composite Concrete-Filled U-Shaped ...

Cyclic Seismic Testing of Composite Concrete-Filled U-ShapedSteel Beam to H-Shaped Column Connections

Cheol-Ho Lee, M.ASCE1; Hong-Gun Park, M.ASCE2; Chang-Hee Park3; Hyeon-Jong Hwang4;Chang-Nam Lee5; Hyoung-Seop Kim6; and Sung-Bae Kim7

Abstract: In this study, the cyclic seismic performance of a concrete-filledU-shaped steel beam toH-shaped steel column connectionswas ex-perimentally evaluated. The concrete-filled U-shaped steel beams were compositely attached to the concrete floor slab. The test was conductedin two stages. The first testing program was carried out on one-sided moment connections to find the most promising connecting scheme.The strengthening scheme, or welding steel plates to the beam bottom flange with minimized stress concentration, was shown to be the mostsatisfactory, and it was used in the second-stage test on two full-scale cruciform specimens. Considering the unique constructional natureof the proposed composite connections, the critical limit states such as weld fracture, local buckling, concrete crushing, and rebar bucklingwere carefully addressed in designing specimens. Test results showed that the connection details and design procedures proposed in thisstudy can successfully control the critical limit states mentioned previously. The proposed connection detail successfully pushed the plastichinging to the tip of the strengthened zone as intended in design, thus effectively protecting the more vulnerable beam-to-column welded joint.The specimens typically exhibited a maximum story drift capacity of more than 5.5% rad, exceeding the minimum limit of 4% rad required ofspecial moment frames. Four of the five specimens tested in this study eventually failed because of a low-cycle fatigue fracture across the beambottom flange at a high story drift greater than 5.0% rad. DOI: 10.1061/(ASCE)ST.1943-541X.0000635. © 2013 American Society of CivilEngineers.

CE Database subject headings: Seismic design; Connections; Composite beams; Columns; Cyclic tests; Frames.

Author keywords: Seismic composite connection; U-shaped composite beam; Cyclic tests; Special moment frames.

Introduction

In the past, the performance of a composite system was often con-sidered to be uncertain and questionable under cyclic seismic loading.However, many studies have steadily demonstrated that well-detailedcomposite members and connections have adequate stiffness,strength, and ductility for seismic design (Leon 1990, 1998; Roeder1998). Currently, the design and construction of composite members

and connections is further evolving in practice. Fig. 1(a) showsa conventional composite beam where a wide flange shape is con-nected to a concretefloor slab via shear studs. In an attempt to enhanceboth structural and constructional efficiency, a new type of compositebeam system that uses a U-shaped steel beam as shown in Fig. 1(b) isproposed in this study. This beam is fabricated through longitudinalwelding of two pieces of cold-formed asymmetric C sections. The U-shaped section may be simply thought of as a closed-section trans-formation of a conventional wide flange open section. However, thissimple change of section configuration results in many structural andconstructional advantages. First, both local and lateral buckling re-sistance of the thin-walled U-shaped section is significantly increasedas a result of concrete filling. Second, the U-shaped section, which isalso used for concrete formwork, enables rapid construction. Third,the thermal sink effect provided by the filled concrete increases fireresistance. Fourth, reducing the beamdepth (or storyheight reduction)is often possible. Last, considering all these, the U-shaped compositebeam can be a less expensive alternative.

Traditionally, most composite beams have been used mainlyfor gravity loading. One of the reasons for this is that compositeconnection details suitable for lateral loading have not beenreadily available to practicing engineers. Compared with the caseof noncomposite steel beam-to-column connections, the seismicdetails for composite connections are difficult to standardize andare seldom covered at length in design codes. In the 2005American Institute of Steel Construction (AISC) Seismic Provi-sions (AISC 2005a) one type semirigid (or partially restrained)composite connection is briefly introduced based on studies byLeon (1990, 1998), but no specific connecting recommendations aregiven for the composite special-moment frames that use H-shapedcolumns. The primary objective of this study was to develop seismic

1Professor, Dept. of Architecture and Architectural Engineering, SeoulNational Univ., Seoul 151-742, Korea. E-mail: [email protected]

2Professor, Dept. of Architecture and Architectural Engineering, SeoulNational Univ., Seoul 151-742, Korea (corresponding author). E-mail:[email protected]

3Graduate Student, Dept. of Architecture and Architectural Engineering,Seoul National Univ., Seoul 151-742, Korea. E-mail: [email protected]

4Graduate Student, Dept. of Architecture and Architectural Engineering,Seoul National Univ., Seoul 151-742, Korea. E-mail: [email protected]

5President, SEN Structural Engineers, Co., Ltd., 121-74 Dangsan-dong6-ga,Yeongdeungpo-gu,Seoul 153-010,Korea.E-mail: [email protected]

6Structural Engineer, SEN Structural Engineers, Co., Ltd., 121-74 Dang-san-dong 6-ga, Yeongdeungpo-gu, Seoul 153-010, Korea. E-mail: [email protected]

7Structural Engineer, SEN Structural Engineers, Co., Ltd., 121-74 Dang-san-dong 6-ga, Yeongdeungpo-gu, Seoul 153-010, Korea.

Note. This manuscript was submitted on September 21, 2011; approvedon April 24, 2012; published online on April 26, 2012. Discussion periodopen until August 1, 2013; separate discussions must be submitted forindividual papers. This paper is part of the Journal of Structural Engi-neering, Vol. 139, No. 3, March 1, 2013. ©ASCE, ISSN 0733-9445/2013/3-360–378/$25.00.

360 / JOURNAL OF STRUCTURAL ENGINEERING © ASCE / MARCH 2013

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http://dx.doi.org/10.1061/(ASCE)ST.1943-541X.0000635

mailto:[email protected]









connection details to connect the U-shaped composite beam to the H-shaped column and evaluate their cyclic seismic performance basedon full-scale cyclic seismic testing.

First Testing Program

The primary objective of this first program was to find the mostpromising connecting scheme among several alternatives. To this end,a one-sided connection test using a stub column seemed sufficient.Because the samedesignmethod is used in the second testing programon cruciform (or two-sided) connections, detailed design consid-erations will be discussed in the relevant sections. Only globalresponses obtained from the test will be presented because of space

limitations. The design information and key test results are sum-marized in Table 1. Fig. 2 shows the connection details. The beamtip (or the point of load application) was assumed as the inflectionpoint. All three specimens, S1, S2, and S3, had nominally identicalconditions except the connection details and shear stud designdescribed here.

Three connecting alternatives through strengthening the con-nection or weakening the beam section were considered to preventpremature fracture of the welded bottom flange, which is one of themost critical limit states that should be controlled. To strengthen theconnection, steel plates werewelded to the bottom and top flanges ofthe steel beam such that the connection strength exceeds the prob-able moment demand at the column face (Fig. 3). In sizing andattaching two bottom reinforcing plates of Specimens S1 and S3,

Table 1. Design Information and Key Test Results of the First Testing Program

General information Specific information S1 S2 S3

Connecting schemes Strengthening theconnection

Applied Applied (heavier) Applied

Weakening the beam Not applicable Not applicable AppliedDegrees of composite action for positive and negativemoments

Fully composite Partially composite forpositive moment (49%)

Fully composite

Members and element sizes U-shaped beam (depth3width3 plate thickness; mm) 4503 2703 6Solid floor slab (length3width3 thickness; mm) 2; 5823 1; 5003 150Rebars (number-diameter) 4-D25 (diameter, 25 mm)

Nominal and measuredmaterial strengths (MPa)

Measured compressive strength of concrete(fck 5 24 assumed in design)

24 25 25

Yield strength of steel platesand rebar

Steel plates (SM490,Fyn 5 325)

446 (6-mm plate from U-shaped beam)390 (35-mm plate from column)

Rebar (SD400, Fyn 5 400) 515 (D25)Flexural strengths of beamand column (kN-m)

Column moment strength Based on nominal strength 2,028Based on measured strength 2,433

Beam plastic moment basedon nominal strength

Positive, M1n 983 757 983

Negative, M2n 2773 2773 2735

Beam plastic moment basedon measured strength

Positive, M1pr 1,285 911 1,294

Negative, M2pr 21,007 21,011 2967

Key test results Maximum connectionstrength (kN-m)

Positive, M1max 1,346 1,345 1,331

Negative, M2max 21,166 21,275 21,146

Cyclic strain: hardeningfactor at assumed plastichinge location

Positive 1.0 1.3 1.0Negative 1.1 1.1 1.1

Maximum plastic rotation (rad) 0.045 0.028 0.035

Fig. 1. Conventional wide flange shape versus U-shaped composite beam: (a) conventional composite beam; (b) U-shaped composite beam

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efforts were made to minimize the notch and to avoid welding on thecold-formed corner of the U-section [Fig. 2(a)]. In Specimen S2,heavier strengthening was used, with the reinforcing plates enlargedinwidth by 40mmand in length by 100mm, respectively [Fig. 2(b)].From the component strength perspective, this reinforcement wasthe strongest among the three, although it was recognized that somenotch and discontinuity at the junction of longitudinal and transversewelds around the reinforcing plates were unavoidable. In SpecimenS3, weakening the beam section was also added around the tip of thestrengthened part by embedding a soft styrofoam into the filledconcrete; thus, a combined strategy was used, with the intention thatthis would more effectively push the plastic hinging away from thecolumn face [Fig. 3(c)]. Lateral ties were provided in all specimensto prevent the buckling of the top rebar.

Shears studs for Specimens S1 and S3 were provided for fullcomposite action for cyclic loads, whereas Specimen S2 wasdesigned for partial (49%) composite action for the positive mo-ment. The aim of the partial composite design of Specimen S2 wasto reduce the tensile force demand on the bottom flange weld underthe positive moment such that a premature tensile fracture of thebottom flange could be avoided more effectively. In this study, the

shear studs were conservatively designed, neglecting the possiblecontribution of the monolithic concrete present between the topflange tips of the U-shape at the beam-slab interface level. Theconcrete is pouredmonolithically by using theU-shape as a formwork.

Cyclic seismic loading according to the 2005 AISC SeismicProvisions (AISC 2005a) (Table 2) was applied to the beam tip usingthe test setup shown in Fig. 4. All specimens had a strong panel zoneand satisfied the strong column-weak beam condition such that thebeam plastic hinge provides all the inelastic rotation. Fig. 5 showsthe hysteretic response and failure mode of each specimen. Thestrengthening scheme successfully pushed the local buckling andplastic hinging away from the column face, thus protecting the morevulnerable welded joint. Specimen S1 [Fig. 5(a)] showed the bestperformance among the three. The connection detail in Specimen 1was used in the subsequent cruciform specimen test. The beamplastic rotation (equivalent plastic story drift capacity) developed inSpecimen 1 was 4.5% rad, far exceeding the minimum limit of 3%rad required of special-moment frames in the AISC Seismic Pro-visions. Specimen 3 also exhibited satisfactory performance (beamplastic rotation of 3.5% rad). However, the additional beamweakening in Specimen 3 [Fig. 5(c)] triggered earlier local buckling

Fig. 2. Connection details tested in the first experimental program: (a) A-A0 section of Specimens S1 and S3; (b) A-A0 section of Specimen S2;(c) longitudinal details


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Fig. 3. Connection strength versus probable moment demand: (a) Specimen S1; (b) Specimen S2; (c) Specimen S3


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and subsequent rapid strength degradation; no benefit was observed.The final failure mode of three specimens was the low cycle fatiguetearing of the bottom flange, which started at the tip of thestrengthened part. Tearing progressed gradually, and completefracture occurred at very high plastic rotation. It is noteworthy that allthe specimens exhibited the plastic positive and negative momentstrengths at the tip of the strengthened zone (see the bottom ofTable 1). Although the bottom flange of Specimen 2 was thestrongest among the three, Specimen 2 showed inferior perform-ances because of the bottom flange fracture, which occurred at 3%

drift level. It should be noted that some notch and discontinuity at thejunction of longitudinal and transverse welds around the reinforcingplates were unavoidable in this specimen. As mentioned before,Specimen S2 was designed for partial (49%) composite action forthe positive moment so that the tensile force demand on the bottomflange weld could be reduced. The partial composite design did notshow any positive effects. As reported in Table 1, significantoverstrength (a cyclic strain-hardening factor of 1.3) for thepositive moment was observed. It appears that this was caused bythe contribution of the monolithic concrete at the interface betweenthe top of theU-shaped beam and the concrete floor slab, whichwasneglected in designing the shear studs. Further study is needed toquantify the contribution of the monolithic concrete to sheartransfer.

In short, Specimen 1 showed the best performance, and theconnection detail in Specimen 1 was used in the subsequent cruci-form specimen test.

Second Testing Program

Basis of Specimen Design

In this testing program, two cruciform specimens (Specimens Aand B) were tested under a two-sided connection configuration. Toevaluate the cyclic seismic performance of the proposed connec-tion detail in more general conditions, additional parameters wereincluded in the second test program as follows. One specimen with

Fig. 4. Test setup for the first testing program

Table 2. Cyclic Seismic Loading According to the 2005 AISC SeismicProvisions (Data from AISC 2005a)

Load step Peak story drift ratio (rad) Number of cycles (n)

1 0.00375 62 0.005 63 0.0075 64 0.01 45 0.0015 26 0.02 27 0.03 28 0.04 29 0.05 210 0.06 211 0.07 212 0.08 2


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a deeper U-shaped section (550 mm deep) was included (the depthof the U-sections in the first testing program was 450 mm deep).The concrete floor slab was formed on the ribbed deck plates in-stead of the flat deck plates used in the component tests. The effectof the panel zone strength on seismic performance was also in-vestigated. Finally, slip-critical bolted splices were included inthe specimen to simulate an as-built condition because the shop-welded beam stubs should be bolt-spliced in the field. The fol-lowing critical limit states were specifically addressed in the

specimen design. Under positive moment rotation, the anticipatedlimit states of fracture of the beam bottom flange, concrete crushingat the floor slab, and the buckling of the rebar in the top of the slabshould be suppressed until the required design ultimate limit statedrift reached. In the negative moment rotation, anchorage failure ofthe welded rebar coupler and local buckling of the U-shaped sec-tion should be the major considerations, among others. Finally, theslippage of the bolted splice when subjected to cyclic loadingshould be prevented.

(a)

(b)

(c)

Fig. 5. Hysteretic response and failure mode of Specimens S1, S2, and S3: (a) Specimen S1; (b) Specimen S2; (c) Specimen S3


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Design of Composite Beam and Connection

Table 3 summarizes the design information and some of the key testresults. Fig. 6 shows the arrangements of slab rebar, shear studs, andlateral ties.As shown inFig. 6, the composite beams are composedofthe U-shaped steel section, a ribbed floor slab, rebar, and shear studs.The depths of the U-shaped steel sections were 450mm for SpecimenAand550mmfor SpecimenB.Madeof thin plates (6 and8mmthick)by cold-forming, Specimens A and B violated the seismic webcompact requirements (AISC 341-05, Section 8.2 b; AISC 2005a) by20 and 10%, respectively. The ribbed metal deck was tag-weldedto the top flange of the U-shaped beam. The floor slab was sized as1,690 mm wide in both specimens, which is equal to the effectiveslab width, or one-fourth of the beam span length (6,760 mm), ofthe specimens according to the AISC Specification (AISC 360-05,Section I3; AISC 2005b). Four (25 mm diameter) longitudinalreinforcing bars were placed in the top of the slab and connected tothe column flange via a welded coupler to assure the negativemoment transfer of the composite section. The shear studswereweldedon the top flanges of the U-shaped section with 115 mm averagespacing for full composite action. The diameters of the studs are 16 and19mm for Specimens A and B, respectively. Both specimens satisfiedthe limitation on the distance from themaximumconcrete compressionfiber to the plastic axis required by the AISC Seismic Provisions (PartII: Section 9.3).

Generally, the capacity design approach underlying currentAISC Seismic Provisions for special-moment frames was fol-lowed. The required flexural strength of the column, shear strengthof the panel zone, and welding strength of the beam-to-columnjoint were calculated by assuming that the plastic hinge will de-

velop at the tip of the strengthened zone as shown in Fig. 7. Asmentioned previously, the strengthening scheme for Specimen S1of the first testing program was used in both specimens. Thestrengthened zone was designed to have greater moment capacitythan the expected moment at the column face calculated by ex-trapolating the expected plastic hinge moment to the column face.The ratios of the moment capacity to the moment demand at thecolumn face were 1.4 (positive moment) and 1.2 (negative mo-ment) for Specimen A and 1.24 (positive moment) and 1.2(negative moment) for Specimen B. In computing the expectedplastic hingemoment, overstrength and cyclic-hardening factors ofthe U-shaped section (SM490 steel) were assumed to be 1.2 and1.1, respectively, according to the Korean Building Code [Ar-chitectural Institute of Korea (AIK) 2009]. The overstrength factorof the rebar (SD400 steel) was taken as 1.25, and the expectedconcrete strength was assumed to be higher than the nominalstrength by 8.0 MPa, according to the American Concrete Institute(ACI) Building Code (ACI 2008).

Column, Panel Zone, and Continuity Plates

Fig. 8 shows the connection details of Specimens A and B. MacRaeet al. (2007) proposed that the overstrength effect of slabs needs to beconsidered in calculating the increased demand on the column andpanel zone. However, the AISC Seismic Provisions does not con-sider this effect. The size of the column was selected to satisfythe strong column-weak beam concept according to the AISCSeismic Provisions (Part I, Section 9.6). Because the column wassufficiently strong compared with the panel zone and the beam

Table 3. Design Information and Key Test Results of the Second Testing Program

General information Specific information A B

Member and element sizes(mm)

U-shaped beam (depth3width3 plate thickness) 4503 2703 6 5503 2703 8Ribbed floor slab (length3width3 thickness) 6; 7603 1; 6903 165Rebars in slab (number-diameter) 8-D25Lateral ties (diameter @spacing) D10@142/88 (average: 115)Shear studs (number-diameter @spacing) 2-f16@142/88

(Average:115)2-f19@142/88(Average:115)

Strengthening plate-bottom flange(number-length3width3 thickness)

2e1503 1003 12 2e2003 1003 15

Strengthening plate-top flange (length3width3 thickness) 1503 1203 6 2503 1203 12Column size (H-depth3width3web thickness3 flange thickness) H-4283 4073 203 35 H-4583 4183 303 50Panel zone doubler plate (length3width3 thickness) 6503 3103 15 7503 3103 15Continuity plate (length3width3 thickness) 3583 1753 15

Nominal and measuredmaterial strengths (MPa)

Measured compressive strength of concrete ( fck 5 24 assumed in design) 30 25Yield strength of steel plates andrebar

Steel plate (SM490, Fyn 5 325) 456 (6-mm plate from U-shaped beam)435(8-mm plate from U-shaped beam)344(20-mm plate from column web)389(30-mm plate from column web)341(35-mm plate from column flange)

Rebar (SD400, Fyn 5 400) 497

Flexural strengths of beamand column (kN-m)

Column moment strength Based on nominal strength 2,028 3,077Based on measured strength 2,128 3,683

Beam plastic moment based onnominal strength

Positive, M1n 1,010 1,667

Negative, M2n 2779 21,199

Beam plastic moment based onmeasured strength

Positive, M1pr 1,379 2,079

Negative, M2pr 21,024 21,521

Column–beam moment ratio 1.50 1.42Panel zone capacity to demand ratio 1.13 1.38

Key test results Maximum connection strength, M1max 1M2

max (kN-m) 2,427 3,517Maximum story drift (rad) 0.054 0.065


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(see the column-moment ratio in Table 3), the majority of the storydrift was expected to be contributed by the beam and the panelzone deformation. The nominal shear strength of the panel zonewas determined by Eq. (1) in the AISC Specification (AISC 360-05, Section J10.6, AISC 2005b). The nominal shear strength de-fined in Eq. (1) presumes some limited panel zone yielding becauseit includes the beneficial contribution of the column flange in thepostelastic range. As shown in Fig. 8, the panel zones of SpecimensA and B were reinforced by the doubler plates because the capacityof the panel zone of the column was weaker than the required shearstrength. The capacity to demand ratios of the reinforced panel

zone were 1.13 for Specimen A and 1.38 for Specimen B, re-spectively (Table 3)

Vpz ¼ 0:6Fycdctw

1 þ 3bct2f

dzdctw

!ð1Þ

where Fyc 5 nominal yield stress of the steel column, dc 5 depth ofthe steel column, dz 5 panel zone depth between the continuityplates, bc 5width of the steel column, tw 5 thickness of the columnweb, and tf 5 thickness of the column flange.

Fig. 6.Arrangement of rebar, studs, and lateral ties for Specimens A and B: (a) slab reinforcement plan; (b) A-A0 section; (c) rebar and lateral ties;(d) B-B0 section; (e) shear studs arrangements


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The 15-mm-thick continuity plates were groove-welded at thelevel of the top rebar and the beam bottom flange based on themaximum expected positive and negative moments (Fig. 8).

Weld and Bolted Splice Design

The welding electrode with minimum Charpy V-Notch (CVN)toughness of 27 J at 229�C was used in the beam-to-columnwelded joint. The welded joints were designed to resist themaximum expected tensile force under the positive and negativemoments. The U-shaped section was first connected to the col-umn flange by all around two-sided fillet welding, and thenstrengthening plates were welded to the bottom and top flange ofthe U-shaped section. The rebar couplers were connected to thecolumn flange by the partial joint penetration groove welding. Asshown in Fig. 6, to prevent buckling of the top rebar, the lateral tiesof 115-mm spacing on average were provided from the columnface to the location, two times the depth of U-shaped beam awayaccording to the special moment frame (SMF) requirement in theACI Building Code (Chapter 21: Section 6.4; ACI 2008). The

bolted beam splice of the column-tree construction, located closeto the beam plastic hinge, is expected to experience significantload reversal under seismic loading. A slip-critically designedbolted splice was included in the specimen design to simulate theas-built condition.

Loading and Measurement Instrumentation

Fig. 9 shows the overall configuration of the cruciform testspecimens. Two actuators in parallel were connected to the topof the column to induce the AISC seismic loading (Table 2).Lateral bracing was provided at the vicinity of the actuator andthe side faces of the floor slab at 1,700mm away from the columncenter line to simulate the lateral bracing effect, which isavailable from the adjacent continuous floor slab in actualbuildings. No lateral bracing was provided at the bottom flangeof the beam because high lateral-torsional buckling resistance ofthe closed U-shaped section was expected. Fig. 10 shows theinstrumentation plan. LVDTs were installed to measure thedeformation of the beam, the column, the panel zone, and therigid body rotation of the test assembly. An extensive set of

Fig. 7. Connection strength versus probable moment demand


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strain gauges was attached to monitor detailed behavior aroundthe connection region.

Test Results

Connection Moment versus Story DriftRatio Relationship

Originally, as shown in Fig. 9, two load cells were attached to thebeam tips to measure the shear force of each beam such that thehysteretic response of each beam could be resolved. Unfortunately,because of a malfunction of the load cells during the test, such data

reduction was impossible. However, the moment strength of theinterior connection can be evaluated as the sum of the positive andnegative moments. The sum of the positive and negative moments atthe column face (M1 1M2) was computed using the statics relationaccording to Eq. (2)

M1 þ M2 ¼ HP�LbL

�ð2Þ

where H 5 column height, L 5 horizontal distance between thebeam supports, P 5 magnitude of lateral loading, and Lb 5 clearspan length of a beam [Fig. 9(a)].

(a)

(b)

(c)

Fig. 8. Connection details of Specimens A and B: (a) connection details; (b) C-C0 section; (c) D-D0 section


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Fig. 11 shows the connection moment (M1 1M2) versus storydrift ratio relationships of two specimens. In terms of the flexuralstrength, the connection moment developed at the column face wasslightly higher (by 1% in Specimen A) or was slightly less (by 2% inSpecimen B) than the actual plastic moment,M1

pr 1M2pr , which was

calculated based on the measured material strengths (Table 3). Bothspecimens showed excellent energy dissipation with limitedpinching and exhibited a maximum story drift capacity of more than5.0% rad, exceeding the minimum limit of 4% rad required ofspecial-moment frames.

(a)

(b)

Fig. 9. Test setup for cruciform specimens: (a) global dimension and test setup; (b) overall view of test setup


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Failure Modes

The photographs in Fig. 12 show the damage and failure modesobserved in Specimens A and B at high story drift levels.

Specimen AAt a story drift ratio of 1.5% rad, initial concrete cracking occurredin the slab. Concrete cracking passing both tips of the columnflanges (transverse to the beam axis) was pronounced. As thestory drift ratio increased from 1.5 to 3.0%, the concrete crackingsteadily increased in the transverse direction. Any visible dam-age was not observed in the beams until a 3% drift level. At a 4%story drift ratio, the bottom flange near the tip of the strengthenedzone started to buckle. It should be noted that, although theU-shaped section was cold-formed from a very thin plate of 6 mmthickness, violating the AISC seismic web compactness re-quirement, the stiffening effects of the filled-concrete success-fully delayed the occurrence of local buckling until a high storydrift level (4% rad). At a 5% story drift, the local buckling in thebottom flange at the plastic hinge location increased further.At 6% story drift, local buckling propagated from the bottomflange to the web. Complete low-cycle fatigue fracture acrossthe bottom flange, which started at the cold-formed radius cor-ner, between the flange and the web, occurred at a 7% storydrift ratio, thus leading to rapid strength degradation of theconnection.

Until the termination of the test, no buckling and frac-ture occurred in the rebar at the top of the concrete slab. Noconcrete crushing was found in the connection region. Plastichinging developed at the tip of the strengthened zone as intendedin the design, and the connection detail proposed was very ef-fective in protecting the more vulnerable beam-to-column weldedjoint.

Specimen BLike Specimen A, initial concrete cracking occurred in the slab ata 1.5% story drift. The observed damage and cracking behavior wasgenerally similar to those of Specimen A until a 4% drift. At a 5%story drift, the steel plate in the bottom web at the plastic hingelocation started to buckle outward. At a 6% story drift, the localbuckling propagated into the middle of the beam web, and then thebottom flange started to buckle. At a 7% story drift, slippage ofthe rebar in the concrete slab occurred along the stud line, and the

bolts in the top flange of the bolted splice started to slip. Theentire web plate at the tip of the strengthened zone buckled signifi-cantly. At a 8.0% story drift, the entire section at the tip ofstrengthened zone buckled, and a gradual fracture of the top spliceplate of the bolted connection was observed at the right beam. Atthe same cycle, a fracture of the top splice plate also occurred in theleft beam, which resulted in rapid strength loss of the connection.No low cycle fatigue fracture of the bottom flange was observed inSpecimen B.

The overall test results of both testing programs indicate that thecritical limit states such as weld fracture, local buckling, andconcrete crushing are suppressed by the proposed connectingscheme and design methods. Most of specimens in this study,except Specimen B, eventually failed because of a low-cycle fa-tigue facture of the beam bottom flange, which initiated from theradius corner of the U-section near the tip of the strengthened part.This may be explained because the cold-formed region such as theradius corner tends to lose its material ductility and fracturetoughness (Hancock et al. 2001).

Story Drift Components

The total story drift is the sum of three components contributed bythe column, panel zone, and the beam [Eq. (3)] is

dtotal ¼ dc þ db þ dp ð3Þ

Fig. 14 shows a comparison of component contributions to the totalstory drift in each drift level. As expected, the column contributionwas negligible because the column was very strong and remainedelastic throughout the response. In both specimens, until a story driftlevel of 3–4%, the contribution of the panel zone increased.However, at a higher story drift, the contribution of the panel zonedecreased, whereas the contribution of the beam increased signif-icantly. At a high story drift more than 5%, the contribution of thepanel zone was very small relative to the beam contribution. InSpecimen A, the maximum panel zone contribution was about 2%,which was achieved at a 4% total story drift [Fig. 14(a)]. InSpecimen B, the maximum panel zone contribution was 1%, whichwas reached at a 3% total story [Fig. 14(b)]. The reason for thedecrease in the panel zone contribution at a large story drift can beexplained as follows. As inelastic shear deformation is imposed

Fig. 10. Measurement plan for the second test


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further on the panel zone after initial shear yielding, the shearstrength of the panel zone continues to increase because of thecyclic hardening of the panel zone. Conversely, the moment ca-pacity of the composite beam unavoidably deteriorates because ofthe local buckling of steel plates, damage of the slab concrete, and/or the slippage of the shear studs. Then the deformation imposedshould be accommodated by the weakened beam.

Strain Responses

Panel ZoneFig. 15 shows the shear strain responses at the center of the panel zone.In Specimen A, the maximum shear strain was 8–10 times the yieldstrain in shear. In Specimen B, the maximum strain was about fivetimes the yield strain in shear. In the shear design of the panel zone, thenominal shear strength of the panel zone was determined

Fig. 11. Connection moment versus story drift ratio relationship: (a) Specimen A; (b) Specimen B


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according to Eq. (1), which permits some limited panel zoneyielding. The second term in the bracket of Eq. (1) is often calledthe column flange contribution (CFC) factor. The shear capacity todemand ratio in the panel zone with neglecting the CFC factor was0.83 for Specimen A and 0.96 for Specimen B, indicating thatpanel zone yielding is the expected behavior. The magnitude ofshear strain recorded in the panel zone indicates that the panel zoneyield was sufficient to develop the additional shear strength fromlocal plastic hinging of the column flanges at each corner of thepanel zone.

RebarFig. 16(a) shows the strain profiles of the rebar around the con-nection. The rebar experienced stable cyclic behavior withoutbuckling and fracture. The strain level of the rebar was the highest at

the column face despite the bottom strengthening. Generally, thestrain levels in Specimen B were lower than those in Specimen A,implying that slippage between the rebar and the concrete slaboccurred in Specimen B, and the composite action was not fullymobilized. The slippage appeared to be the primary reason why themaximum moment developed in Specimen B was slightly less thanthe plastic moment calculated by using the measured materialstrengths. This unexpected slippage in Specimen Bwas suspected tobe caused by a few factors, such as the reduced slab cross-sectionalarea as a result of the presence of the ribbedmetal deck, congestionof studs and rebar, and insufficient curing of the concrete. Theconcrete quality for SpecimenBwas inferior. As shown in Table 1,themeasured concrete strength of Specimen B barely exceeded thenominal strength, whereas that of Specimen A showed com-pressive strength higher than the nominal strength by 25%. Asa result, the cohesive forces in Specimen B for slip resistance

Fig. 12.Failuremodes around connection region (SpecimenA): (a) local buckling initiation at 4.0%; (b) top slab concrete cracking at 3.0%; (c) top slabconcrete cracking at 5.0%; (d) local buckling at 5.0%; (e) local buckling at 6.0%; (f) top slab concrete cracking at 6.0%; (g) low-cycle fatigue fracture at7.0% (left beam); (h) low-cycle fatigue fracture at 7.0% (right beam)


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between the rebar and the concrete might not be as sufficientlydeveloped as with Specimen A.

Beam Bottom FlangeFig. 16(b) shows the maximum strain levels of the beam bottom flangenear the connection and the bolted splice plates in the bottom flange. Inboth specimens, the strain levels at the strengthened zone remainedelastic. The strain levels of the beam bottom flange at the assumedplastic location (or just outside of the strengthened zone) reached morethan 10 times the tensile yield strain of the steel.

Summary and Conclusions

In this study, cyclic seismic testing of a composite concrete-filled U-shaped steel beam to H-shaped column connections was conducted.

On the basis of the test results, the following conclusions can bedrawn:(1) Overall, the critical limit states in the composite connections

such as weld fracture, local buckling, and concrete crushingwere well controlled by the connection details and designmethods proposed in this study.

(2) The connection detail proposed per the strengthening strategysuccessfully moved plastic hinging to the tip of the strength-ened zone as intended in the design and exhibited the storydrift capacity far exceeding the minimum limit required ofspecialmoment frames according to the currentAISC seismicprovisions.

(3) Four of the five specimens in this study eventually failedbecause of a low-cycle fatigue facture of the beambottom flange that initiated from the cold-formed corner

Fig. 13. Failure modes around connection region (Specimen B): (a) top concrete cracking at 4.0%; (b) local buckling initiation at 5.0%; (c) top slabconcrete cracking at 6.0%; (d) local buckling at 6.0%; (e) slip at splice at 7.0%; (f) local buckling at 7.0%; (g) local buckling of entire section at 8.0%;(h) fracture at splice at 8.0%


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of the U-section near the tip of the strengthened part.However, tearing of the steel plate progressed gradually,and a complete fracture occurred at a high drift levelof more than 5% rad. Although slippage between theconcrete slab and rebar occurred in one of the two cruciformspecimens, this was not critically detrimental to theoverall seismic performance. No lateral bracing was pro-vided to the beam bottom flange. However, lateral insta-bility was not observed in any specimen during test becauseof the closed-section nature of the U-shaped compositebeam.

(4) Considering all these test results, it can be concluded that theproposed connection can develop consistent and reliable

seismic performance sufficient for use in composite special-moment frames. The proposed connectionmay not be a low-damage solution following a severe earthquake, whichwould require replacement of a damaged beam, but isexpected to resist moderate to low earthquakes with negli-gible damage.

Acknowledgments

Financial support for this study by the Small and Medium BusinessAdministration of South Korea (Grant No. 0420-20100066) andCreative Global Technology Development for Sustainable Con-struction is gratefully acknowledged.

Fig. 14. Comparison of story drift components: (a) Specimen A; (b) Specimen B


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Fig. 15. Shear strain response at the center of panel zone: (a) Specimen A; (b) Specimen B


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Fig. 16. Maximum strain responses of rebar, beam bottom flanges, and splice plates: (a) rebar; (b) beam bottom flanges and splice plates


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References

American Concrete Institute (ACI). (2008). “Building code require-ments for structural concrete and commentary.” ACI 318-08,Detroit.

American Institute of Steel Construction (AISC). (2005a). “Seis-mic provisions for structural steel buildings.” AISC 341-05,Chicago.

American Institute of Steel Construction (AISC). (2005b). “Specification forstructural steel buildings.” AISC 360-05, Chicago.

Architectural Institute of Korea (AIK). (2009). “Korean building code andcommentary.” KBC 2009, Seoul.

Hancock, G. J., Murray. T., and Ellifrit, D. S. (2001). Cold-formed steelstructures to the AISI specification (civil and environmental engineer-ing), Marcel Dekker, New York.

Leon, R. T. (1990). “Semi-rigid composite construction.” J. Construct. SteelRes., 15(1–2), 99–120.

Leon, R. T. (1998). “Composite connections, structural engineering andmaterials.” Construct. Res. Commun., 1(2), 159–169.

MacRae, G., Clifton, G. C., and Mago, N. (2007). “Overstrength effects ofslabs on demands on steel moment frames.” Proc., Pacific StructuralSteel Conf., Heavy Engineering Research Association (HERA), Auck-land, New Zealand, 275–282.

Roeder, C. W. (1998). “Overview of hybrid and composite systems forseismic design in the United States.” Eng. Structures, 20(4–6), 355–363.


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http://dx.doi.org/10.1016/0143-974X(90)90044-H

http://dx.doi.org/10.1016/0143-974X(90)90044-H

http://dx.doi.org/10.1016/S0141-0296(97)00035-7

Cyclic Seismic Testing of Composite Concrete-Filled U-Shaped ...

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