CIA Robust-Buildings-in-Australia.pdf

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Concrete Solutions 09 Paper 5b-2

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Robust Buildings in Australia

Colin GurleyTAFENSW Sydney Institute of Technology, Ultimo 2007 [email protected]

Synopsis: This paper grew out of a study-tour to Chicago in April 2006, a few months afterpublication of the NIST Study on the 911 attack on the World Trade Center. The five events precedent

to this paper were: (1)Accidental domestic kitchen gas-stove explosion at Ronan Point, UK 1968destroyed a 22-storey precast concrete apartment building; (2)The 1995 terrorist truck-bomb attack onthe Murrah Federal Building in Oklahoma City; (3)The 2001 terrorist attack on the World Trade CenterTowers, New York; (4)Fire 2005 35-storey Windsor Building, Madrid. Zero fatalities!; and (5) Fire 200550-storey Government Building, Venezuela. Zero fatalities!This paper deals with three issues: (1) Fire: is it an issue for structural engineers and should buildingsbe designed to burn-out without collapse even if sprinklers fail? (2) Ductile detailing: how earthquakedetailing can drastically improve resistance to accidental/terrorist explosions. (3) Lost column analysis:a column, lost to accidental explosion, terrorist attack or earthquake, should not lead to collapse that isdisproportionate/progressive.Australian standards have mentioned robustness since soon after the Ronan Point event. These rules,currently at AS/NZS1170.0s6 are bare-bones statements of principle. The present purpose is to putsome flesh on them.

Keywords: Ronan Point, World Trade Center, Murrah Building Oklahoma, fire, ductility, earthquake

Fire at the World Trade Center 2001

Fire following the attack at the World Trade Center has been described in FEMA403/ASCE(1),NIST2005(2), in Gurley(3,4,5) and in the Extended Abstract to this paper. NIST seems clear that:

Dislodgement of the thermal insulation (SFRM) in the debris path of the aircraft impact was anessential condition for the collapses of both towers:

o The WTC towers would likely not have collapsed under the combined effects of aircraftimpact damage and the extensive, multistory fires that were encountered on September11, 2001 if the thermal insulation had not been widely dislodged or had been only

minimally dislodged by aircraft impact. NIST2005, pp. 176, 185 Dislodgement of thermal insulation in the debris path at the impact floors was caused by the

aluminium shrapnel debris created by the impact:o Wider dislodgement at other floors above and below the impact zone probably occurred

due to the shaking caused by the impact buto The shrapnel dislodgement in the debris path of impact zone (several storeys) was, of

itself alone, sufficient to cause collapse when combined with the simultaneous structuraldamage and the multistory fires NIST2005, p. 178 ; and

The physical condition of the insulation before impact was not an important issue:o Some localized substandard thicknesses including zero thickness caused by locally

careless application and/or dislodgement during adjacent building maintenance may haveexisted but did not substantially affect the outcome.

NIST recommended the following far-reaching changes:

Recommendation 8. NIST recommends that the fire resistance of structures be enhanced byrequiring a performance objective that uncontrolled building fires result in burnout without partial orglobal total collapse. Such a provision should recognize that sprinklers could be compromised,nonoperational, or nonexistent NIST2005, page 211 ; and

Recommendation 9. NIST recommends the development of: (1) performance-based standardsand code provisions, as an alternative to current prescriptive design methods, to enable thedesign and retrofit of structures to resist real building fire conditions, including their ability toachieve the performance objective of burnout without structural or local floor collapse and (2) thetools, guidelines, and test methods necessary to evaluate the fire performance of the structure asa whole system. Standards development organizations, including the American Institute of Steel

Construction, have already begun developing performance-based provisions to consider theeffects of fire in structural design NIST2005, p. 211


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NIST does not believe that buildings should be designed for aircraft impact NIST2005, p. 216 . NISTseems to suggest that, for future projects, burnout should be demonstrated by a structuraldisproportionate collapse analysis for fire perhaps including lost column elements. That structuralanalysis has to assume that sprinklers will have failed. The sprinkler assumption does help to simplifyissues by separating structural responsibilities from those related to the sprinkler system.

Reliability of SFRMs will be crucial; SFRMs must work even if, particularly if, spr inklers fail.

Structural analysis for fire is still in its infancy and the NIST investigation has had to cobble togetherthree separate software packages:

The Fire Dynamics Simulator can predict the room temperatures and heat release rate values forcomplex fires to within 20% when the building geometry, fire ventilation, and combustibles areproperly described NIST2005, p. 184 ;

The Fire Structure Interface, developed for this investigation, mapped the fire-generatedtemperature and thermal radiation fields onto and through layered structural materials to within theaccuracy of the fire-generated fields and the thermophysical data for the structural componentsNIST2005, p. 184 ; and

Structural analysis software ANSYS 8.0 finite-element program.

It will take some time for an integrated software package to cover all of this and unresolved researchissues may emerge in the process. In the meantime it seems appropriate to focus on the fire-protection of structural elements: core walls and steel beams and faade columns.

Fire Rating of Structural Connections

The provisions that were used for the WTC towers did not require specification of a fire-ratingrequirement for connections separate from those for the connected elements. The Investigation Teamwas unable to determine the fire rating of a connection where the connected elements had differentfire ratings, and whether the applied insulation achieved that rating. NIST2005, p. 198

Sprayed Fire-Resisting Materials (SFRMs)

Recommendation 6. NIST recommends the development of criteria, test methods, and standards: (1)for the in-service performance of SFRMs also commonly referred to as fireproofing or insulation usedto protect structural components; and (2) to ensure that these materials, as installed, conform toconditions in tests used to establish the fire resistance rating of components,assemblies, and systems. This should include the following:

1. Improved criteria and testing methodology for the performance and durability of SFRM e.g.,adhesion, cohesion, abrasion, and impact resistance under in-service exposure conditions e.g.,temperature, humidity, vibration, impact, with/without primer paint on steel for use in acceptanceand quality control; and

2. Inspection procedures, including measurement techniques and practical conformance criteria forSFRM in both the building codes and fire codes for use after installation, renovation, or

modification of all mechanical and electrical systems and by fire inspectors over the life of thebuilding. NIST2005, pp. 210, 211.

There is discussion on SFRM reliability in America now including an article in the New York Times 9September 2008. The perceived issue is not just whether SFRMs are successful in laboratory test-furnaces but whether they will be robust enough to resist dislodgement during any abnormal event.Clearly the SFRM materials used at WTC were not sufficiently robust to resist dislodgement during the2001 attack and this did lead directly to the collapses. Alas NISTIR7563 of 2009 (6), the draft BestPractice Guidelines does not seem to have addressed this issue yet.

Even though NIST does not believe that buildings should be designed for aircraft impact, it wouldnevertheless be relevant and interesting to see the outcome of tests simulating aircraft attack by astorm of aluminium fragments sprayed, as debris shrapnel at aircraft speeds, at comparable structural

elements of steel-framed buildings and concrete-framed buildings. Would 20 mm, or more, concretecover resist dislodgement more effectively than current SFRMs? This author is not aware that anysuch tests have been done.


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NIST on Core Walls

The functional integrity and survivability of the stairwells was affected by the separation of thestairwells and the structural integrity of stairwell enclosures. The shaft enclosures were fire rated butwere not required to have structural integrity under typical accidental loadsthere were numerousreports of stairwells obstructed by fallen debris from damaged enclosures. NIST2005, p. 177

Analysis indicated that the aircraft impact rupture of large return air shafts and related ductworkcreated a major path for vertical smoke spread in the towers. NIST2005, p. 188

At the time of the design and construction of the WTC towers, there were no explicit minimumstructural integrity provisions for the means of egress stairwells and elevator shafts in the building corethat were critical to life safety. The building core, generally designed to be part of the vertical gravityload-carrying system of the structure, need not be part of the lateral load-carrying system of thestructure In the case of the WTC towers, the core had 2 h fire-rated, gypsum partition walls with littlestructural integrity, and the core framing was required to carry only gravity loads. Had there been aminimum structural integrity requirement to satisfy normal building and fire safety considerations, it isconceivable that the damage to stairways, especially at the floors of impact, may have been lessextensive. NIST2005, p. 195

Recommendation 18. NIST recommends that egress systems be designed: 2 to maintain theirfunctional integrity and survivability under foreseeable building-specific or large-scale emergencies...1. Within a safety-based design hierarchy that should be developed, highest priority should beassigned to maintain the functional integrity, survivability, and remoteness of egress components andactive fire protection systems sprinklers, standpipes, associated water supply, fire alarms, and smokemanagement systems NIST2005 p. 216 ;2. The design, functional integrity, and survivability of the egress and other life safety systems e.g.,stairwell and elevator shafts and active fire protection systems should be enhanced by consideringaccidental structural loads such as those induced by overpressures e.g., gas explosions , impacts, ormajor hurricanes and earthquakes, in addition to fire separation requirements. In selected buildings,structural loads due to other risks such as those due to terrorism may need to be considered. WhileNIST does not believe that buildings should be designed for aircraft impact, as the last line of defensefor life safety, the stairwells and elevator shafts individually, or the core if these egress components

are contained within the core, should have adequate structural integrity to withstand accidentalstructural loads and anticipated risks; and3. Stairwell remoteness requirements should be met by a physical separation of the stairwells thatprovide a barrier to both fire and accidental structural loads.... NIST 2005 p. 216

It seems obvious that load-bearing concrete cores, as usually used in Australian practice with 200minimum thick walls and N12 minimum rebar each way each face, would amply provide a minimumstructural integrity requirement to satisfy normal building and fire safety considerations, such that it isconceivable that the damage to stairways, especially at the floors of impact, may have been lessextensive. NIST2005, p. 195. It is difficult to think of any alternative construction or material capableof comparable performance. Concrete cores will also better protect active fire services, risers forwater and for alarm/communication systems and reduce the vertical spread of smoke to higher stories.

Ductile detailing of strong-column/weak-beam build ings

Earthquake engineering was re-invented from 1972, in New Zealand and in California, because ofdisappointing performance in the Alaska earthquake of 1964 and the Los Angeles (San Fernando)earthquake of 1971. Even so, there were then still occasional articles from well-meaning (and quiteeminent) American engineers suggesting that a soft bottom story was a good thing!

This authors 1972 memory is that SEAOC had already adopted a strong-column/weak-beamprovision to the effect that, at every beam/column joint, the total column bending strength (above +below) must be 20% larger than the total beam bending strength (left + right). This was certainly amajor step in the right direction. It may have been simplistic in the sense that the important thingwas/is to prevent soft-storey failures in which all of the post-elastic strain energy is concentrated intoone or a small number of storeys. A more advanced approach looks at storey strength rather than

individual columns and seeks sufficient strong-columns to prevent a soft-storey failure. Of coursemost shear-walls are (very) strong columns in one or both directions so the use of a shear-wall/corealmost automatically prevents soft-storey failures. Note however that special ductile detailing of weakcolumns is essential so that they do continue to carry vertical loads.


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Nevertheless this paper will focus on strong-column/weak-beam buildings in which each column isstronger than the immediately adjacent beams. This is relevant because (see Extended Abstract)ASCE has concluded that ductile detailing would have increased the cost of the Oklahoma Building by1 2% and reduced the extent of collapse by 80%.

Collapse mechanisms for moment-resisting frames

Figs 1 to 3 show the usual three skeletal collapse mechanisms for beams in multi-storey buildings.The plastic hinges are indicated by black circles:

Sag = Tension-bottom yield

Hog = Tension-top yield

All of the plastic hinges are in the beams rather than columns as discussed in strong-column/weak-beam buildings above.

The present author has written elsewhere (7,8) about how these mechanisms can be modified withdogleg hinges so as to provide for the exact rigid-plastic yield-line design of reinforcement inconcrete beams for bending and for shear.

Figure 1Gravity load only

Figure 2Wind/Earthquake-dominated beam with reduced gravity load

Figure 3Gravity-dominated beam with wind/earthquake

Bottom rebar at columns and other supports

The first and most important issue in ductile detailing is that of continuity of bottom rebars at supports,particularly intermediate supports. This is discussed in more detail in Gurley(3).

For lost column situations, the first question is whether there is some stronger structure at higherlevels as there clearly was at the World Trade Center (the hat trusses). On the other hand, there aremany buildings which approximate a typical floor building in which every floor needs to support itself.This is discussed in much more detail in USGSA2003(9) and in Gurley(10,11).

At each floor above a lost column the span will have doubled and locations directly above the lostcolumn will now be mid-span regions. If there is no beam bottom rebar at each floor above the lostcolumn then there will an un-reinforced section with zero strength and zero ductility.


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AS3600:2001 requires only that 25% of mid-span bottom rebars continue some distance (say 50 mm)past the face of support unless, of course there is a calculated nett positive moment under the designlateral load combination. The corresponding clause of ACI 318 draws a distinction for beams that arepart of a primary lateral load resisting system for which bottom rebars should be fully anchoredbeyond the column face regardless of the theoretical design moment combination whether tension-topor tension-bottom.

This writer believes that this matter reflects a long-standing oversight in AS3600.

If the moment caused by, say, the 500-year wind as the design wind load is less than the momentcaused by 90% of the dead load then there need be no effective bottom rebars at columns. But the10,000-year wind load may be up to 150% of the 500-year wind. What then? There is no tension-bottom ductility if there is no effective bottom rebar! And this for a pulsating load?

The situation with earthquake is much worse. The 10,000-year earthquake may be 290% (NZS1170.5)of the 500-year earthquake but the designer of a frame of ordinary ductility can still reduce the designmoment to just 38% (1/2.6) of the 500-year value. Thus the 10,000-year earthquake moment can be7.5 (2.9*2.6) times the gravity load (0.9G) moment. And still there need be no effective bottom rebar!And again there will be no ductility for an un-reinforced section especially under alternating load. And

perhaps the ratio 10,000-years/500-years is greater in Australia because the 500-year figures arealready so low in this intra-plate region.

Finally one should note that bottom-rebars at supports can also act as compression reinforcement ifproperly restrained from buckling (with ties) thereby improving ductility under tension-top moments.

Special duct ile design for beams of ductile frames

The AS/NZS1170 gravity load for Fig 1 is (1.2G+ 1.5Q) where the live load Qmay be subject to anarea reduction but not a duration reduction. The stress-resultants, moment and shear-force are those

determined by the above load with whatever plastic redistribution (< 30% for ku< 0.20) the

designer selects. A designer can use redistribution (< 30%) to maximize the extent of designrepetition within a project in order to reduce design and construction costs. The design is subject tothe usual reliability (strength reduction) factors: (bending = 0.80; shear = 0.70) .

The gravity loads for Figs 2 and 3 are OR(0.9G,G+cQ) and the, say, 500-year earthquake is

(Eu) . Redistribution, (< 30%) , and reliability factors are then applied. The selection of longitudinal

rebar carried is carried out but after that the design approach changes. The moment capacities arecalculated for the as-to-be-drawn longitudinal rebar using an over-strength factor in the range

(longitudinal over-strength = 1.25 to 1.5) depending on the code. The over-strength factor is

meant to include rebar over-strength as-supplied but also the effects of strain-hardening. The gravity

load is maintained at OR(0.9G,G+cQ) but the earthquake load is varied so that a mechanism

occurs. Shear reinforcement is then designed with a reliability factor (= 0.70 - 1.0) againdepending on the code.

Longitudinal rebar in special ductile beams

There seems to be general agreement that the minimum reinforcement in beams produce a bendingstrength with a small (say 20%) margin over the un-reinforced cracking moment. In AS3600:2001, thisminimum only applies at critical sections. Elsewhere, subject to bond and anchorage, it can be just afraction of that. The major earthquake codes want to apply this minimum content at all points, top andbottom along the length of the beam presumably on the reasoning that a plastic hinge can occuranywhere under earthquake load.

This writer has always preferred a content of tension rebar leading to a neutral axis depth parameterku 0.20 with, where appropriate, allowance for compression reinforcement. This is, indeed,

somewhat conservative. He is of the view that larger beams, lightly reinforced are more economic than


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tiny beams, crammed with rebar that requires steel-fixers to emulate watch-makers.

The longitudinal rebar:

At midspan bottom will be about half of the top rebar at supports At supports bottom again about half of the top rebar at supports At midspan top will be about the minimum content described aboveLaps are prohibited within a support, within (2d) of the ends of a beam or within (2d) of any mid-span

plastic hinge as in Fig 3. In general the top splice will be towards mid-span from the quarter-spanposition and the bottom splice towards the support from the quarter-span position. There is not muchroom so one has to do the best one reasonably can.

The maximum diameter of longitudinal rebars is the parallel support/column dimension/20, that is 20maximum rebars through a 400 column. These bars are yielding in tension on one column face andheavily stressed or yielding in compression on the other and this does imply higher bond-stresses thanthe clauses on bond and anchorage.

Shear reinforcement details

ACI-318-Ch21 defines a hoop as a closed tie or a spiral. In either case the ties may be made ofseveral pieces so long as every piece terminates with a 145 or 180 seismic hook. 90 degree hooks

were permitted by AS3600, as one remembers, in the 1980s but are now illegal because the covershell concrete outside the tie spalls under overload and 90 degree hooks are then un-anchored. Allhooks on ties and stirrups must now bend through 135 or 180 degrees so that they are anchoredinside the confined core concrete. ACI-318-Ch21 does permit 90 degree hooks at one end only ofinterior cross-ties which must alternate position on successive cross-ties; this is a concession to steel-fixing problems.

Shear reinforcement in beams

Under ACI-318 Ch 21, closed hoops are required:

At laps Within (2d) of the support face

Within (2d) of any mid-span hinge as in Fig 3.The maximum spacing in the zones above is:

MIN(d/ 4,8* dbarlongitudinal

,24* dbarhoop

,100 at lap-splices, 300 elsewhere)

One cannot imagine that one needs spacings < 100 even if d < 400 but that is what the ACI code

Ch21 requires. Elsewhere stirrups (not necessarily hoops) are spaced at maximum (d/ 2) ; although

there may not be much significant elsewhere. For wider beams, every second rebar across the widthshould have a cross-tie at a maximum cross-section spacing of 350.

This shear reinforcement should also be sufficient for the shear strength described above. There is noACI-318-Ch21 minimum content so that falls back on the general provisions for shear strength. Thiswriter has written elsewhere about design for shear (7,8). He believes that the AS3600 minimumsmeared yield-strength of 0.35 MPa is too low.

Columns in moment-resisting f rames

Columns are divided into 2 zones:

Confined zones are at each end for a length = MAX D,b,Ln

/ 6,450( )but the full height if thecolumn is not 20% stronger in bending strength than the directly attached beams. The NZS3101strength enhancement is quite a lot more than 20%.

The lesser confined remainder.

Shear reinforcement for confined zones should be closed hoops which, in each direction provide asmeared tensile yield strength as a confining pressure of:

MAX 0.09 fc,0.03 f

c

Ag

Ach

1)

=

MAX 2.25,0.75Ag

Ach

1)

MPa for fc= 25 MPa


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where Ag= Gross area of column andA

ch= Confined area of column measured to outside of ties.

Note that these values are much larger than the current AS3600 figure of 0.35 MPa for minimumshear reinforcement. There were similar provisions in past editions of AS3600 at a time when AS3600recognized that spirally reinforced circular columns were inherently superior to tied columns. Indeedthe strength of the confined core of such columns were then considered to be increased by a term

4.1p where p= Confining pressure provided by spiral reinforcement. To the best of this authors

knowledge, this provision was correct but was lost in some code rewrite.

There is no suggestion in ACI-318-Ch21 that beam hinge zones be similarly confined but this authorsmemory is that there may be in NZS3101.

The maximum spacing for shear reinforcement in confined zones is:

MIN D / 4,b / 4,6dblo ng itudinal

,100 for cross-tie spacing = 350, 150 for cross-tie spacing 200( )and that elsewhere is: MIN 6dbar

longitudinal,150( )

Beam-column join ts in duct ile frames

The shear-strength of beam-column joints is still a major issue. The beam longitudinal bars areexpected to exceed yield-strength on one face in tension and on the opposite face in compression.Column size shall be >= 20 times the diameter of longitudinal beam rebars in the same direction.Confinement reinforcement ranges down to half of that mentioned above for confined zones. The factthat columns may face a biaxial attack from diagonal earthquakes does not seem to have beenadequately considered; how can beams that are driving the shear force into the column adequatelyconfine it?

Beam-column joints are open to theoretical study using this authors dogleg hinge method for thecollapse-mechanism solution of plane-stress problems (7,8).

Ductile shear-walls/cores

Many Australian concrete cores approximate thin-walled RHS (rectangular hollow sections) withnumerous internal walls and numerous openings crossed by coupling beams.

Columns in tall buildings are not normally ductile in the sense, say, that the neutral axis depth ratio

ku< 0.20 under the maximum abnormal event co-incident vertical gravity load. Shear walls/cores

can and, arguably should be ductile in that same sense. This does imply that each leg of theperimeter wall of the RHS, acting as a compression flange under different lateral load-directions,should be able to carry most of the maximum abnormal event gravity load (not the full design gravityload) plus the bending compression force for all of the rebar elsewhere acting in bending tension. Thismay be a controversial view but this author does not see how one can otherwise claim a full ductilityreduction for a shear core when the neutral axis depth is much deeper. This may involve increasingthe thickness of RHS perimeter walls to 300 - 400 or more. It does not help to increase the

compression-rebar in such flanges because this will increase the area of rebar in tension when thelateral load reverses. This increased thickness can be limited, say to the height of a plastic hinge upfrom the base. At the base of a tall building, there is often some spare space in the vertical air ductsthat can be used for this purpose.

The coupling-beams over openings including doors to stairs and lifts will usually fall within the ACI-

318-Ch21 definition of deep coupling-beams as L /D< 4 . In this regard ACI-318Ch21 followsNZS3101 and requires tied-column cages on both diagonals. This writer is quite certain that this willprovide ductility in many situations that may be otherwise have been rather brittle. The author is notyet certain that this is always necessary in Australia; he has written on this and does intend to updatethose writings (13).

The shear-strength of thin-wall RHS sections with multiple parallel webs is one that has barely beenmentioned in the literature. It is an aspect that this author does mean to address. Shear-walls are, inthe first instance, just large cantilever beams and the first objective is that they yield in flexure ratherthan shear. There will be many squat walls in low-rise buildings for which that may be impossible and


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many other tall buildings in which the shear is shared over 3 or 4 parallel webs.

Lost column Analysis

Lost column analysis is required under UK Approved Document A3 on Disproportionate Collapse (14)and USGSA2003 (9). The purpose is to prove that a disproportionate/progressive collapse will notoccur if a column is removed by an accidental explosion or by a terrorist attack.

In important cases, this will require a non-linear finite element analysis. However this authorsbimoment method can be used for quick hand calculations of yield-line bending strength excludingcatenary action (10,11,12). This seems appropriate particularly for corner columns.

Conclusions

This paper has considered three issues relating to the robustness of buildings: fire, ductile detailing,and lost column analysis.

For fire, the major conclusions for structural engineers follow from the NIST recommendations that:

Structural engineers take responsibility for structural collapse includingcollapse during fire and That buildings should be designed to burn-out without collapse even if sprinklers fail and

That fire may coincide with a lost column event where columns are lost to accidental explosions,to terrorist-attacks or to earthquake.

That SFRMs need regular inspection, once a new standard has emerged, to ensure that they arelikely to survive abnormal events in a condition good enough to do their job.

For existing buildings, such an inspection could take place whenever a floor is vacated with someupper limit, say at 25 years.

Ductile detailing determines the ability of a building to respond to an abnormal event in a benign way.It is an issue quite separate from (and arguably more important than) the determination of designlateral load. This writer has long believed that some minimum standard of ductile detailing should bedetermined by the Building Importance as defined in BCA. He reached that conclusion whileinspecting the damage from the San Fernando earthquake 1971 and he is encouraged that theFEMA227/ASCE 1996 report (15) on the Oklahoma City bombing reaches similar conclusions.

Lost column analysis seeks to show that the loss of a column whether by accidental explosion,terrorist bomb or earthquake will not cause a collapse that is disproportionate or progressive.Analytical methods include:

Non-linear finite element analysis and The bimoment method which does yield-line analysis by hand of the slab including intervening

beams treated as torsion-free grillages.(10,11,12)

References

1. FEMA403/ASCE May 2002: World Trade Center Building Performance Study. Federal EmergencyManagement Agency, Washington

2. NIST2005 . Collapse of the World Trade Center Towers. Final Rep., National Institute ofStandards & Technology, NIST NCSTAR1, Washington, D.C.

3. Gurley, C.R. Protecting life and reducing damage in earthquakes and terrorist attacks. AEES06Conference: Earthquake Engineering in Australia, Canberra, 24-26 November 2006.Downloadable from

4. Gurley, C.R. Progressive Collapse and Earthquake Resistance. Practice Periodical on StructuralDesign and Construction, ASCE Virginia, February 2008. Republished in Concrete in Australia,December 2008

5. Gurley, C.R. Structural Design for Fire in Tall Buildings. Practice Periodical on Structural Designand Construction, ASCE Virginia, May 2008.

6. NISTIR 7563: 2009. DRAFT for Public Comments. Best Practice Guidelines for Structural FireResistance Design of Concrete and Steel Buildings. US Department of Commerce. Feb 2009.

7. Gurley, C.R. 2008. Plastic shear strength of continuous reinforced beams. New Zealand Society

for Earthquake Engineering Conference, Wairakei April 2008.8. Gurley, C.R. Collapse-Mechanism Design of Ordinary Concrete Beams for Shear. Submitted for

publication May 2009.


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9. USGSA June 2003 Progressive Collapse Analysis and Design Guidelines for New Federal OfficeBuildings, General Services Administration, Office of the Chief Architect, Washington.

10. Gurley, C.R. Concrete plasticity in structural design practice. Magazine of Concrete Research,Thomas Telford Ltd, ICE London. October 2008.

11. Gurley, C.R. Plastic disproportionate collapse at lost corner columns. Magazine of ConcreteResearch, Thomas Telford Ltd, ICE London. February 2009.

12. Gurley, C.R. Plastic Yield-line Analysis of Torsion-Free Flat Slabs. ASEC2008, Melbourne June2008. Extended and republished Australian Journal of Structural Engineering, Online 2009.

13. Gurley, C.R. Core Coupling-Beams in Tall Buildings AEES07 Conference, Wollongong 2007.Downloadable from www.aees.org.au

14. Approved Document A 2004 A3 Disproportionate Collapse; The Building Regulations 2000,London. Downloadable from http://www.planningportal.gov.uk/ Click on Professional User; Clickon Building Regulations; Click on Technical Guidance; Click on Part A. OR GO TO:http://www.planningportal.gov.uk/england/professionals/en/4000000000067.html

15. FEMA227/ASCE August 1996: The Oklahoma City Bombing: Improving Building PerformanceThrough Multi-Hazard Mitigation. Federal Emergency Management Agency, Washington.

Dedication

This paper is dedicated to the memory of Prof Tom Paulay of Canterbury University, New Zealand.

Tom died in Christchurch on Sunday 28 June 2009.

Tom was the leading research expert world-wide on ductile shear-walls and coupling-beams; seeAppendix. Tom also had well-told recollections of life as a Hungarian cavalry officer, riding a real 4-legged horse and dodging Russian tanks in WW2.

Tom and this author first met in Christchurch in July 1972 when Tom was the critical path expert for aproject there for which this author was a structural designer. We recently (2008) lunched with Tom inone of those buildings. It is a shear-wall building.

Appendix on earthquake engineering ducti le frames and shear-walls

Lateral load resistance can be achieved with moment-resisting frames, with vertical steel bracing, withshear-walls/cores or with any combination thereof. In the view of this writer, the earthquakeengineering literature and the majority of buildings in New Zealand and, maybe, California, have longbeen disproportionately interested in ductile moment-resisting frames.

Tall buildings in Sydney have had load-bearing concrete shear-cores since the 1960s; perhapsinfluenced by the extensive prior experience of the Australian building industry with large slip-formedsilos at ports and numerous rail-sidings in wheat-growing regions.

The Los Angeles (San Fernando) earthquake of 1971 seemed to confirm this writers own Sydney-based prejudice:

Ductile-frame buildings suffered severe non-structural damage and were vacant for many months(even years) under repair whereas

Shear-wall buildings were mostly re-occupied within a few days. These effects are most important when all/most of the buildings in a community have sufferedsimilar fates so that daily lives are disrupted, cash-flow from most ductile-frame buildings isstopped and the repair industry has to rely on more distant sources of labour and materials.

Even after the San Fernando earthquake, most buildings in New Zealand and California continued tobe designed as ductile-frame buildings. Perhaps engineers were overly pre-occupied with ductility andtended to overlook the advantage of shear-wall stiffness and strength particularly in terms of non-structural damage. There was also (and may still be) a conviction that shear-walls were, necessarily,less ductile than moment-resisting frames. Indeed this authors 1972 memory is that even if oneprovided a very strong/stiff shear-wall which could carry all of the lateral load, one was neverthelesscompelled to provide ductile moment resisting frames to carry a minimum 25% of that same lateralload. This may have been a good thing. Understanding of three-dimensional action was then limited

because of the lack of three-dimensional software and even 2-dimensional analysis required computeraccess by modem from Australia to an American main-frame computer. The warping torsionalstrength of open shear-wall sections is small and faade moment frames are a very effective way tosupplement torsional strength.


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Another interesting comparison is that the Australian engineering literature talks about shear-coreswhile the New Zealand and, to a lesser degree, Californian literature talks about shear-walls. This isnot merely a linguistic difference. The main shear-core at Centrepoint Sydney, directly supportingSydney Tower was designed and built 1967 1971 and has, from memory, about 15 discrete cells,mostly rectangular and each several metres in plan extent. This core is a complex, closed multi-cell,thin-walled box beam with huge torsional strength. It is the largest of 4 cores in Centrepoint. The sizeand complexity of a core necessarily depends on the services provided and hence on the area of eachtypical floor and the height of the building. Centrepoint has an unusually large typical floor area butonly about 15 floors served by 8 lifts and a proportionate area of vertical air-ducts. The core alsoincorporates fire-stairs, toilets and lunch rooms.

Later Australian buildings are several times taller but have rather smaller floor areas. There are more,perhaps many more, lifts but rather less vertical air ducts because 30-storey and taller buildings useintermediate plant-rooms. 1972 New Zealand shear-walls, when used at all, were smaller and simplerwith few, if any, closed cells and only warping torsional-strength. Often some walls enclosing liftswere non-structural masonry which raises the risk of seismic collapse of the masonry into the lift-wells.

Colin GurleyCol Gurley graduated BE (Civil, Hons) from the University of Sydney and MEngSc (Concrete

Structures) from the University of NSW. He practised as a structural designer in southeast Australiaand New Zealand with Wargon Chapman (now Hyder), with John Grill (Worley Parsons) and withKinhill (KBR). His career highlights include periods as the lead structural design engineer for the tallestbuildings in Sydney and Adelaide, structural design of offshore oil platforms, and the NZIE (IPENZ)Hume Prize 1982 for work on the earthquake resistance of heritage buildings.

Col has retired from routine practice, but retains an active interest in areas such as the robustness andsafety of tall buildings, in lessons to be learnt from the collapse of the World Trade Center, earthquakeengineering, and the study of structural plasticity and the quest for exact and appropriate collapse-mechanisms for yield-line analysis of slabs loaded out-of-plane and plane-stress problems forwalls/plates loaded in-plane.

CIA Robust-Buildings-in-Australia.pdf

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