73_ftp

8/14/2019 73_ftp

1/11

73 2012 Ernst & Sohn Verlag fr Architektur und technische Wissenschaften GmbH & Co. KG, Berlin Structural Concrete 13 (2012), No. 2

This paper examines the problems of the seismic design of

precast concrete structures as indicated by the effects of the

LAquila earthquake of 2009. The behaviour of such structures on

that occasion is analysed on the basis of detailed surveys per-

formed on site on a relevant number of buildings just after the

earthquake.Following this analysis, which points out the inadequacy of

the cladding panel connections, the paper presents a systematic

definition of the problem, examining the behaviour of the whole

system of structure + panels. Proposals are made for the typology

of one-storey buildings for industrial uses, some possible alterna-

tive solutions that ensure the stability of all the construction ele-

ments. Specific calculations for a typical precast building give

the order of magnitude of the forces and/or displacements for

which the connections are to be designed.

Keywords: precast structures, seismic behaviour, structural connections

1 Introduction

The violent shake that struck the town of LAquila and itsneighbourhood on the morning of 6 April 2009 had a de-structive power of 5.8 on the Richter scale and a momentmagnitude of 6.3 [1]. From the data of the National Seis-mic Network in different zones, there were horizontalpeak ground accelerations between 0.35 and 0.40 g (0.60 gnear the fault) which, from the seismic design point ofview, correspond to the maximum seismicity expected inItaly. There were also particularly high vertical peakground accelerations, of the order of magnitude of 0.50 gin some zones.

This violent earthquake struck a stock of hundredsof different precast constructions spread over some 10zones around LAquila. Most of them were industrial one-storey buildings and in a few cases commercial buildingsof two or three storeys. The territory struck by the earth-quake was classified for a seismic action of about 0.25 g.Some buildings were designed with the 1996 Italian Code,which already gave some specific detailing rules for seis-mic behaviour. Many others were designed with an-

tecedent codes which lacked such rules. Obviously, thelatest Italian Code updated to 2008 did not have any in-fluence on the earthquake effects and we can wonderwhether an anticipated issue of this code would have hada positive influence.

The first thing to be noted is the general good behav-iour of the structures on their own, together with the not-so-good behaviour of wall claddings, with the failure ofmany panels. On the one hand, these results confirm thereliability of the seismic design of precast frame structureswith hinged beams, as codified after more than 15 years oftheoretical and experimental research. On the other hand,they show the inadequacy of the present design approachto fastening systems for perimeter wall panels.

As on preceding occasions, the earthquake has pro-vided a dramatic lesson that has to be learned properly.We remember, for example, the lesson learned from the

Friuli earthquake of 1976: the fall of a beam from its bear-ing had shown that under seismic action it is not possibleto rely on friction for the transfer of horizontal forces. Thecombination of contemporary horizontal and verticalshakes can reduce the gravity action on which friction isbased. This fact had been learned by the National ItalianAssociation of Precast Concrete Producers, whichpromptly circulated a specific recommendation to itsmembers. From that initiative, a general correct designpractice was derived for mechanical connections betweenbeams and columns which later led to the specific rule inthe latest version of Eurocode 8.

For the LAquila earthquake of 2009, the effects

should again be examined and indications derived for up-dating design criteria and execution rules. The workshould confirm the things revealed as adequate and leadto improvements to things revealed as inadequate, propos-ing also propose possible new alternative design rulesbased on innovative approaches.

Very important to this end will be the results of anEuropean research on the seismic behaviour of connec-tions in precast structures involving 16 partners from theEuropean countries mostly subjected to seismic risks, in-cluding the principal centres of seismic experimentation.The work started in 2009, will last three years and lead to

the synthesis of the experiences gathered over more than15 years of experimental and theoretical research on thesubject of precast construction by the different nationaland international schools [2], [3] involved.

Articles

Precast concrete structures: the lessonslearned from the LAquila earthquake

Giandomenico Toniolo*

Antonella Colombo

DOI: 10.1002/suco.201100052

* Corresponding author: [email protected]

Submitted for review: 25 October 2011Revised: 31 January 2012Accepted for publication: 14 March 2012

8/14/2019 73_ftp

2/11

2 Observations after the earthquake

During the month following the earthquake, the authorsmade visits to the area of LAquila, including the industri-al zones of Poggio Picenze, Monticchio, Ocre, Varranoni,Bazzano, Pila, Pettino and Popoli in particular. Some 50buildings were examined with different degrees of depthdepending on access possibilities, the availability of specif-ic information and the interest in the case: from simple ex-ternal visual inspections to detailed inspections of the in-ternal and external parts of the building.

The visits have been addressed on the basis of the in-dications of some local operators in the sector who hadexperienced the earthquake personally and knew the situ-ation in the area well. The same operators guided us onmany visits, obtained permission for access and provideda great deal of technical information about the construc-tions of which often they were the designers and produc-ers. This allowed us to cover all the situations where rele-vant damage had been noted.

Fig. 1 shows an emblematic image: a recently con-

structed building where the structure (columns, beams,roof elements) has essentially retained its integrity, but awhole wall of vertical panels has collapsed. The cause of

74

G. Toniolo/A. Colombo Precast concrete structures: the lessons learned from the LAquila earthquake

Structural Concrete 13 (2012), No. 2

the collapse is shown in Fig. 2: the panel fastenings failedunder forces not considered in the design.

A list of the principal serious damage to the struc-tures is presented first: a shed beam that has fallen from its bearing (together

with the adjacent roof span) a long-span roof element that has fallen from its bearing the overturning of some TT floor elements (at erection

stage) some spalling of column edges some spalling of beam-to-column bearings

In addition, there are different minor types of damage,such as cracking in columns, local spalling of the edges atthe bearings of floors on beams and of beams on columns.The large displacements during the earthquake also led todifferent damages to finishings such as internal partitions.

3 Structural damage

Fig. 3 shows the shed beam that has fallen from its bear-

ing. The beam dragged down the adjacent span of the roofwith it, which at the time of the picture had already beenremoved. The building is a cowshed in Fossa.

The part of the structure where the collapse oc-curred was built in 1997. The shed beams were placed inpocket supports at the tops of the columns. Supports andbeams were connected with a through-dowel. Under theseismic shakes, this connection did not work, probably be-cause of the absence of a mortar infilling to fix the dowel.

Fig. 4 shows the roof from which a long-span ele-ment fell (the other elements were removed after the earth-quake because they were considered unsafe). The building

is a vehicle workshop in Pile built in 1991. The ends of theroof elements were placed on specially shaped supportingelements and fixed with steel dowels. Under the seismicshakes, the edges of the supporting elements broke aroundthe dowels (Fig. 5) because of insufficient anchorage tocope with the forces. Again, friction did not contribute ad-equately.

These are the only two instances of structural ele-ments in finished precast buildings collapsing aroundLAquila. But in the Fossa cowshed, serious damage oc-Fig. 1. Cladding panel collapse in Ocre

Fig 2. Detail of fastening failure Fig. 3. Shed beam failure in a cowshed in Fossa

8/14/2019 73_ftp

3/11

75



curred in a part of the structure built in 1994. At the top ofthe columns, the lateral walls of some pocket supportsbroke (Fig. 6) pointing out the importance of adequatelateral restraint against the overturning of beams. Some

spalling occurred at mid-height (where the longitudinal re-inforcement is curtailed), with buckling of the longitudi-nal bars (Fig. 7) because of an excessive spacing of the stir-rups with respect to the flexural ductility demand underearthquake loads. This type of damage occurred in someother buildings, too.

Concerning the collapses of structural elements,Fig. 8 shows the case of a building under construction in

Monticchio where the cladding panels also dragged downtwo TT floor elements to which they were connected. Thestructure was not finished: if the planned concrete top-ping had been added to the floor, the collapse would nothave occurred. Floor elements not yet fixed to the struc-ture also fell in a few other buildings under construction.

The two-storey building in Monticchio used as agymnasium received a particularly strong shake with avery high vertical component. Fig. 9 shows a close-up of aseriously damaged bearing for a beam. No collapses oc-curred, but the building had much other relevant damage.Local failures of bearings also occurred in some other

buildings, e.g. the factory in Varranoni shown in Fig. 10.Minor damage not affecting the resistance of thebuildings was more numerous. Small cracks can be seen atthe bases of several columns, but also at mid-height, pre-

Fig. 4. Collapse of roof element in Pile

Fig. 6. Failure of pocket support

Fig. 5. Edge spalling of supporting element

Fig. 7. Buckling of longitudinal bars

Fig. 8. Failure of wall panels and floor elements Monticchio

8/14/2019 73_ftp

4/11

sumably corresponding with the curtailment of the longi-tudinal reinforcement. Also frequent was spalling at thecorners of supporting elements under concentrated bear-ing actions due to floor elements. Finally, signs of largedisplacements have been noticed, up to 150 mm at thetop of some buildings. This has a relevant impact on theconsiderations presented in the following section regard-

ing the design of wall cladding systems.

4 Damage to wall claddings

Collapses of wall panels caused by the failure of their fas-tenings were more numerous. These collapses concerneda relevant number of existing buildings (around 15 %).The forces received under the earthquake were muchhigher than those calculated in the design stage on the ba-sis of local behaviour.

The effects are those shown in Fig. 1 mentioned ear-lier, which refers to a recently constructed building in

Ocre. The collapsed wall was in the NW-SE direction ofthe strongest acceleration component. The fastenings,made from channel bars, have been forced in the tangen-tial transverse direction for which they were not designed.

76



The anchorage head worked as a lever, pulling out theedges of the channel (Fig. 2). In the orthogonal walls, theforce, normal to the connection plane and related to thelocal mass of the panels and not to the whole mass of theroof, had sufficient resistance.

Such collapses occurred in different buildings, in-volving both vertical and horizontal panels, as shown in

Figs. 11 and 12 respectively (buildings in Bazzano andMonticchio).

Failures occurred not only with channel bars, but al-so with other types of fastenings (Figs. 13 and 14). It is nota question of product inadequacy, but of inadequate de-sign of the connection.

Fig. 15 shows the interior of a factory in Poggio Pi-cenze: the structure of columns, beams and roof elementssuffered no damage, but a whole row of horizontalcladding panels fell down. The cause of the collapse isshown in Fig. 16, i.e. the failure of the fastenings. Fig. 17shows the strengthening promptly applied to the remain-

ing panels after the earthquake in the form of strong bolt-ed steel angles.At the conclusion of this brief survey made in the

field, it should be pointed out that the heavy damages pre-

Fig. 9. Serious damage at a bearing in Monticchio

Fig. 10. Spalling at a bearing in Varranoni Fig. 12. Failure of a horizontal panel in Monticchio

Fig. 11. Detachment of a vertical panel in Bazzano

8/14/2019 73_ftp

5/11

77



sented above concern only a minority of buildings. Thelarge majority of precast structures passed the earthquakecheck without relevant damage.

5 Design criteria

Present design practice for the precast structures under

consideration is based on a bare frame model where theperipheral cladding panels are entered as masses onlywithout any stiffness. The panels are then connected tothe structure with fixed fastenings designed with a localcalculation on the basis of their mass for anchorage forcesorthogonal to the plane of the panels.

This approach is not sound the LAquila earth-quake has demonstrated this. Panels fixed to the structurein this way come to be an integral part of the resisting sys-tem, conditioning its seismic response as that of a dualwall-frame system with lower energy dissipation capacity.The high stiffness of this resisting system leads to much

higher forces than those calculated from the frame model.These forces are related to the global mass of the floorsand are primarily directed in the plane of the walls. Theunforeseen intensity and direction of the forces caused

many fastenings to fail, leaving the frame of columns andbeams practically undamaged.The old criterion leaving the walls to break since af-

ter their failure the remaining structure will resist by itself

Fig. 13. Broken fastening

Fig. 14. Wrenched bracket

Fig. 15. Factory in Poggio Picenze with failed panels

Fig. 16. Failure of channel bars in the factory

Fig. 17. Connection strengthened with steel angles

8/14/2019 73_ftp

6/11

anyway does not work when failure implies the fall ofpanels with weights of up to 10 t. The mortal danger ofthese collapses requires a different approach. And theseconsiderations hold true for all precast and cast-in-situconcrete structures: the fall of a large masonry claddingrepresents an equally serious danger.

From these observations, three possible solutions forthe cladding system can be presented, the principles ofwhich are described below.

A first solution, here called isostatic, is based onwall panels connected to the structure with supports thatallow the free development of the large displacements ex-pected for the frame structure under seismic action. Thissolution allows us to pursue the traditional design ap-proach of the analysis applied to a bare frame model.

A second solution, here called integrated, is basedon wall panels connected to the structure with an ipersta-tic arrangement of fixed supports that make them part ofthe overall resisting system. This solution requires a com-pletely new design approach with an analysis applied to adual frame-wall model that involves unexplored participa-

tions of the roof diaphragms.Both these solutions require the specific design of

new connectors: these must allow the large free displace-ments of the isostatic solution, or transfer the high forcesexpected in the integrated solution.

A third solution can be mentioned as proposed bysome operators in the sector, based on the present prac-tice of under-proportioned fixed connectors, expectingthem to fail under seismic actions, but providing the wallpanels with a back-up system made of elementary devicesthat intervene in the case of incipient collapse. This solu-tion, here called secured, could be mainly applied to the

seismic retrofitting of many existing structures.

5.1 Isostatic wall system

Different panel-to-structure connection systems can en-sure an isostatic arrangement of the panels. Examples ofsome of these systems are presented below with referenceto both vertical and horizontal panels, pointing out someproblems of practical application.

Fig. 18a shows a vertical panel referred to an orthog-onal axis system, where x is oriented horizontally in thepanel plane,y is oriented orthogonally to that plane and z

78



is oriented vertically, parallel to the gravity force. The ori-gin is placed in a corner at the base side of the panel.

Four connections are placed at the corners of thepanel, indicated byA, B, CandD. These connections areintended to provide only translational restraint withoutany rotational restraint. The possible joint connectionswith the adjacent panels are indicated by E and F. Usual-ly, connections A and B are attached to the foundationbeam, connections CandD to the top beam.

The same reference system is applied in Fig. 18b to ahorizontal panel, for which connectionsA, B, CandD areusually attached to the columns and E and Frefer to thepossible joint connections with adjacent panels, where theuncertain friction effect due to the weight of the superim-posed panels may act.

Tables 1 and 2 indicate the effect of the supports inthe three directionsx, y and z for the vertical and horizon-tal panels respectively, where the symbols are: f = fixed i = indifferent s = sliding

0 = absent

Strictly speaking, for an isostatic arrangement in the planeof the panel, the indifferent supports should be defined assliding supports but, in expectation of the negligible ef-fects of the small linear deformations of the panel, theycan be executed with the simpler fixed connections. How-ever, this choice has to be verified for connections Cand

D of the vertical panel because it leads the panel to col-laborates in the vertical bearing of the loads applied aftercompleting the connections.

In the plane of the vertical panel (in thex direction),

the connection system defined in Table 1 ensures horizon-tal displacements of the structural frame independent ofthe panel, which remains supported at the base, on thefoundation beam, providing its stability by itself (Fig. 19).In the orthogonal y direction, the panel remains simplysupported at its upper and lower ends, following, withoutreactions, the vibratory motion of the structure, to whichit transfers the inertia force due to part of its mass.

Choosing the simpler type of connection for the in-different supports, i.e. the fixed one, the system requirestwo types of connectors, one with full support and onewith partial support that allows one of the three trans-lations; this is in expectation of large displacements up to

Fig. 18. Schemes for vertical (a) and horizontal (b) panels

Table 1. Supports for the vertical panel

A B C D E F

x f i s s 0 0

y f f f f 0 0

z f f i i 0 0

Table 2. Supports for the horizontal panel

A B C D E F

x f i s s s sy f f f f 0 0

z f f i i 0 0

8/14/2019 73_ftp

7/11

79



15 cm. It should be noted that the base connectionsAandB at the foundation beam cannot be simple bearings,but must be mechanical connections with a bilateral ef-fect.

In the plane of the horizontal panel (in the x direc-tion) the connection system defined in Table 2 binds the

displacements of the panels to those of the joints with thecolumns on which the inertia forces due to their mass aretransferred (Fig. 20). For a free motion without reactions,it is essential that no relevant friction arises at the jointwith the superimposed panels, and this requires the inter-position of a free space or adequate unusual deformableseals. Due to the friction, it is therefore more difficult withthe horizontal panels to create an isostatic system thatdoes not react with the structure.

In the y direction, the horizontal panels follow thevibratory motion of the columns without reactions, trans-ferring the inertia forces due to their mass to the four

joints.Fig. 21 shows another possible solution with an iso-static pendulum arrangement of vertical panels, connect-ed at the base and the top with connections placed in the

middle of the horizontal sides. In this arrangement, thecladding system can freely follow the motion of the roof,displaying relative sliding db/h at the joints of the adja-cent panels, where d = top horizontal displacement, b =

panel width and h = height of upper connection. Onlyfixed connections are required, unless the z support is re-leased to avoid vertical actions on the panels due to thegravity loads transmitted from the connected roof ele-ment.

It should be noticed that, in their rocking motionaround the base support, the adjacent panels, besides slid-ing tangentially with respect to each other, get closer by anamount that, for the expected small values of the topstorey drift, is lower by one order of magnitude than thetop displacement and can be easily offset by ordinary jointplay. Equally negligible is the lowering of the top connec-tion.

The connections of the corner panels, which had dif-ferent shortcomings during the LAquila earthquake, haveto be investigated in more depth.

5.2 Integrated wall system

The same pendulum support system can be transformedinto an integrated system if the E and Fconnections areadded to prevent the relative sliding of the panels at theside joints (Fig. 22). Particular fasteners have been de-signed and tested for these joint connections in order todissipate energy under a violent earthquake [4], [5].

The ordinary arrangement of the connections placedin the four corners of the panels, which make them collab-orate with the structure, is based on full translational sup-ports and the columnsA, B, CandD of Tables 1 and 2 all

Fig.19. Isostatic supports for vertical panels

Fig. 20. Isostatic supports for horizontal panels

Fig. 21. Isostatic pendulum system of connections

Fig. 22. Iperstatic system with interconnected panels

Fig. 23. Iperstatic connection system for vertical panels

8/14/2019 73_ftp

8/11

become filled with f. Fig. 23 shows this arrangement forvertical panels, where the distortion effects at the topbeam can be seen. These effects may be avoided if onlyone connection is placed on the top beam, as indicated inFig. 24. In this solution, the panels behave in their planelike cantilevers fixed at the base and placed side by side.

Similar solutions can be used for horizontal panels:one with four connections as described in Fig. 25 leadingto distortion effects at the columns; another without dis-tortion effects where connectionsA andB at the columnsare replaced by the connections E on the underlying panel(Fig. 26).

The technological solution of fully fixed connectionsis simplified, but the integrated participation of the panels,with their stiffness, in the vibratory motion of the structur-al system leads to very high forces at the connections. In

80



terms of present production, new products with improvedcapacities are needed. Of course, it is not only the steelconnector that needs improvements, but also the nearbyconcrete part of the panel that has to resist without antic-ipated failures. Furthermore, not only the calculation ofthe connections has to be updated, but also the calcula-tion of the roof diaphragm through which the inertiaforces are transmitted to the resisting lateral walls.

Anyway, sizing the connections shall be carried outcase by case, with a global analysis of the structural systemand quantifying the force and displacement distributions.With a much higher structural stiffness, small displace-ments are expected and the serviceability limit state re-quirements (limitation of damage) are easily satisfied.However, more complex resistance and ductility verifica-tions at the no-collapse limit state are expected, addressedmore to the walls than to the columns.

The resistance and plastic deformation capacities ofthe panels are to be quantified for in-plane actions. As-suming low-ductility connections, a capacity design calcu-lation should be set up and tested for the over-proportion-

ing of the connections with respect to the ultimatestrength of the panels in their structural arrangement. Fol-lowing these investigations, the value of the behaviour fac-tor q (force reduction factor) must also be defined for thisstructural system, which does not seem comparable withwhat is defined as a wall system in the design codes be-cause of the great difference in the detailing and the sup-ports.

5.3 Secured wall systems

Finally, we shall introduce as a small seed to germinate

Fig. 27, which refers to the third solution, i.e. a back-upline of elementary fasteners securing against the fall ofpanels added to a structural system designed following thetraditional practice of under-proportioned connections.The theme belongs to the wide scope of the assessmentand possible retrofitting of the building stock: a very com-plex problem that is difficult to solve and for which only afew technical suggestions can be advanced here.

Fig. 24. Alternative Iperstatic system for vertical panels

Fig. 25. Iperstatic connection system for horizontal panels

Fig. 26. Alternative iperstatic system for horizontal panels Fig. 27. Device securing against falling

8/14/2019 73_ftp

9/11

81



6 Analysis of a typical building

The order of magnitude of the displacements and forcesthat are demanded by the isostatic and the integratedsolutions respectively as described above is deduced be-low. As a typical building to be examined, the one-storeyprecast industrial building illustrated in Figs. 28, 29 and30 (50 41 m on plan) has been chosen. The columnshave a square section of 60 60 cm, are 5.2 m high andare placed at a spacing of 10 m along three longitudinallines. The prestressed concrete beams (with a constant

depth of 1.2 m) are positioned at the top of the columnsalong these lines. Resting on the beams are the TT pre-stressed concrete roof elements, each 70 cm deep, 2 mwide and 20.5 m long. A strip 1 m wide is left open for askylight every two elements. Vertical wall panels 2.5 mwide are placed along the perimeter to complete thebuilding.

Concrete C45/55 is used for beams and roof ele-ments, C30/37 for columns and cladding panels. Thecladding panels are sandwich wall elements with an inter-mediate insulating layer. The amount of reinforcement is

Fig. 28. Plan of the typical building

Fig. 30. Transverse section

Fig. 29. Longitudinal section

8/14/2019 73_ftp

10/11

not relevant for the present analysis, which is aimed onlyat quantifying displacements and forces in the panel-to-structure connections.

In a first case, this building is assumed to be standingby itself with cladding panels on the four perimeter sidesin a doubly symmetrical arrangement. In a second case, itis intended to be one of the two parts of a double lengthbuilding divided by a seismic joint and in this case the wallpanels are on three sides only, with an asymmetricarrangement in one direction.

The analyses are applied to a three-dimensionalmodel that reproduces closely the geometry of the struc-ture together with the distribution of the masses and of theelastic stiffnesses of the construction elements, includingthe in-plane stiffness of the roof elements. The arrange-ment of the supports consists of fully fixed supports at thecolumn bases, hinged beam-to-column connections in thehorizontal and vertical planes of the beam, fixed beam-to-column connections in the orthogonal vertical plane, andspherical hinges between the beams and the two ribs ofthe roof elements.

In the calculation model, the wall panels are repre-sented by plate elements connected to the foundation bytwo spherical hinges, whereas at the top they are connect-ed to the beam by two joint elements, the stiffness ofwhich is assumed to be very low or very high each time inorder to simulate the two quoted solutions of isostatic andintegrated connections. These arrangements are those de-scribed in Figs. 18a and 19 and Table 1.

The masses involved in the vibratory structural re-sponse are those of the seismic action combination withonly permanent loads at their characteristic or nominalvalues (with G= 1.0). These loads are evaluated automat-

ically by the computer program: 4.0 kN/m2

distributed onthe roof and 3.0 kN/m2 distributed on the peripheral wallcladdings. No unintended eccentricities of the masseshave been introduced; obviously, the systematic eccentric-ities due to the asymmetric wall arrangement remain.

The seismic action corresponds to a reference peakground acceleration ag = 0.25 g, with the response spec-trum of Eurocode 8 for a soil categoryB (with S = 1.2).The value q = 3.5 has been assumed for the behaviour fac-tor. This assumption seems to be consistent with the bareframe structure (isostatic connection system) for columnproportioning that takes into account the storey drift limi-tation at the serviceability limit state and with the dual

wall-frame structure (integrated connection system) torepresent its reduced dissipative capacities.

Dynamic modal analyses have been performed withthe help of a computer program for defining the principalvibration modes, subsequently evaluating forces and dis-placements separately for longitudinal (x direction) andtransverse action (y direction). Figs. 31 and 32 show thethree-dimensional representation of the model for thesymmetric and the asymmetric arrangements respectively.

7 Results of the analysis

For the symmetrical wall arrangement, the first two modescorrespond to pure translations alongy andx, with natur-al vibration periods for the isostatic connection system of1.03 and 0.80 s respectively (common for this type of

82



frame structure). A third significant vibration mode corre-sponds to a torsional rotation with a natural vibration pe-riod of 0.74 s. These periods are much lower for the inte-grated connection system (0.65, 0.42 and 0.41 s) becauseof the much higher stiffness of the dual frame-wall struc-tural system.

For the asymmetric wall arrangement with isostaticconnection system, the first three modes remain practical-ly the same as for the symmetric arrangement since theframe structure, disconnected from the walls, remainssymmetrical. The different mass distribution modifies thenatural vibration periods a little, to 1.03, 0.79 and 0.74 s.For the integrated system, the asymmetry of the wall sub-stantially modifies the response, with a first translational-rotational mode along y that keeps a high natural vibra-tion period of 0.86 s because of the absence of thestiffening wall on one side, a torsional mode that becomesthe second mode with a low natural vibration period of0.46 s and a symmetrical translational mode along x with

a natural vibration period of 0.41 s.For connection design purposes, we shall consider the

maximum values of the behaviour parameters of the re-sponse that are for the isostatic support system the rela-tive horizontal displacements between panels and structureat their connections, and for the integrated support systemthe horizontal components of the forces transmitted be-tween panels and structure at their connections. The struc-ture analysed can be assumed to be representative of the ty-pology, so the maximum values quoted can be taken asindicative of the order of magnitude of the parameters.

These parameters are given in Table 3 together with

some other data regarding the overall behaviour of thestructures. The terms max slide and max force indicatethe displacement and force in the connection quoted above,max drift indicates the maximum top displacement of the

Fig. 31. 3D model for symmetric arrangement

Fig. 32. 3D model for asymmetric arrangement

8/14/2019 73_ftp

11/11

83



structure, and max shear and max mom indicate thecomponents of the internal force at the column base.

It must be remembered that the calculation of the dis-placements and forces has been performed using a modaldynamic analysis with linear elastic behaviour of the ele-ments and a design response spectrum reduced by a factorq = 3.5 for a peak ground acceleration equal to ag =1.2 0.25 g = 0.30 g, which corresponds to the no-collapse

limit state. The computed values of the displacements havetherefore been amplified by the same factor q = 3.5 to ac-count for the elasto-plastic non-linear behaviour of thestructure.

As can be seen in Table 3, the maximum relative dis-placement between panels and structure for the isostaticsupport arrangement is 7.0 cm. The maximum force trans-mitted between panels and structure for the integratedarrangement is 104 kN.

For the symmetrical structure in the longitudinaldirection, the integrated support system practically halvesthe internal forces at the column base, transferring most of

the action to the lateral walls because of their much higherstiffness.

8 Conclusions

The survey and calculations reported above give the firstindications for the necessary updating of the design crite-ria for the connection system for the wall cladding panelsof precast structures.

Introducing a proper margin with respect to the onlyexample treated in the text (e.g. for higher seismic ac-tions), 1.75 7.0 12 cm indicates the free drift capaci-ty of the panel fastenings for the isostatic solution to the

connection system, and 1.75 10.4 18 t indicates theforce transmission capacity of the panel fastenings for theintegrated solution.

This last strength capacity relates to the single con-nection as a whole not only the steel fastener, but alsothe adjacent concrete parts of the connected elements. Areduction in the required strength is possible if joined byan adequate plastic deformation capacity or with an addi-tional system of connections between the panels [6].

Some problems are still pending with reference to theseismic design of the structural assembly in the case of anintegrated panel connection system that involves the inter-

action between the spatial frame of beams and columns, thebox wall cladding system and the roof diaphragm.This presentation is a long list of subjects to be inves-

tigated in greater depth and will require some years of re-

search. The way to reach a complete knowledge of theseismic behaviour of the constructions is still long. Andthis is valid not only for precast structures. The LAquilaearthquake in particular was a dramatic reminder of theinadequacy of cladding and partition walls, both masonryand precast, for all types of construction. This is a topic ofprimary importance for safeguarding lives when earth-quakes occur.

Acknowledgements

The present work was performed within the scope of theSAFECAST research project supported by the contribu-tion of the European Commission in the FP7-SME-2007-2Programme with Grant agreement No. 218417 of 2009.

References

1. Menegotto, M.: Experiences from LAquila 2009 earthquake,

Proceedings of the 3rd fib Congress, Washington D.C., 2009.2. Colombo, A., Negro, P.: Tailoring experimental strategy and

set-up: the long story of the seismic behaviour of precast struc-tures, 3AESE, San Francisco, 2009.

3. Biondini, F., Toniolo, G.: Experimental research on seismic

behaviour of precast structures, Industria Italiana del Cemen-to No. 3, 2009.

4. Iqbal, A., Pampanin, S., Buchanan, A., Palermo, A.: Improved

seismic performance of LVL post-tensioned walls coupledwith UFP devices, 8th Pacific Conference on Earthquake En-

gineering, Singapore, 2007.

5. Shultz, A. E., Magana, R. A., Tadros, M. K., Huo, X.: Experi-mental study of joint connections in precast concrete walls,

Proceedings of 5th national Conference of Earthquake Engi-neering, 1994.

6. Biondini, F., Dal Lago, B., Toniolo, G.: Seismic behaviour of

precast structures with dissipative connections of claddingwall panels, ANIDIS Congress, Bari, 2011.

Dr. Antonella Colombo

Assobeton

Milan, Italy

Email: [email protected]

Prof. Giandomenico Toniolo

Politecnico di Milano

Ingegneria strutturale

p.za Leonardo da Vinci 32

Milan 20133, Italy

Email: [email protected]

Table 3. Maximum values of behaviour parameters

Symmetric arrangement Asymmetric arrangement

along x along y along x along y

isostatic integrated isostatic integrated isostatic integrated isostatic integrated

max slide (mm) 56 70 55 70

max force (kN) 104 83 100 73

max drift (mm) 59 31 80 59 67 35 73 49max shear (kN) 62 31 47 38 62 31 47 44

max mom (kNm) 362 191 331 256 358 189 331 306

73_ftp

Documents

Transcript of 73_ftp